Energy Dissipating Façade Systems Designed to Reduce ...

Energy Dissipating Façade Systems Designed to Reduce Structural Response during

Earthquakes

A thesis submitted in fulfilment

of the requirements for the degree of

Doctor of Philosophy

By

Pouya Abtahi

Centre for Infrastructure Engineering

Western Sydney University

November 2017

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ii

ACKNOWLEDGMENTS

I am very grateful and indebted to my Principal Supervisor Prof Bijan Samali for his

confidence in me and dedication to my education, for his unlimited support and

supervision and encouragement throughout the course of this study. His insight and

excellent suggestions were extremely important in finalizing this thesis. I also want to

thank Dr Ali Saleh as my co supervisor for his contributions to this thesis. I would like

to thank Western Sydney University (Institute for Infrastructure Engineering) and

University of Technology, Sydney (Centre for Built Infrastructure Research) for

providing Postgraduate Research Scholarship to carry out my research project. I would

also like to thank Dr Guido Lori from Permasteelisa Group for his continual support,

and excellent technical support. He has taught me not only the fundamentals that I

sought but also the strategic problem solving that are critical to their application.

I would also like to thank Permasteelisa Group for its financial support and their

managers, Dr Danijel Mocibob and Dr Marc Zobec for the management of this ARC

Linkage research project. I wish to also extend my thanks to fellow postgraduate

students and friends for their support and contributions to this research with whom I

shared the ups and downs of completing this research project. Finally, I wish to

express my gratitude to my wife Marsa, parents Javad and Nahid and my younger

brother Pirouz who so generously supported me, and encouraged me to focus on this

work. And to whom this thesis is dedicated.

iii

LIST OF PUBLICATIONS RELATED TO

THIS THESIS

Journal Articles & Conference Papers

1. Evaluation of the effect of smart façade systems in reducing dynamic response of

structures subjected to seismic loads, B Samali, P Abtahi, “Journal of Earthquakes

and Structures” August 2016.

2. Performance of flexible façade systems in reducing the lateral displacement of

concrete frames subjected to seismic loads, P Abtahi, B Samali, M. Zobec, T. Ngo,

22th ACMSM: “Materials to Structures: Advancement through Innovation”,

Sydney, Australia December 2012.

3. Evaluation of Effect of Sacrificial Bracket Façade Elements in Reduction of

Dynamic Behaviour of Concrete Structural Models during Seismic Activities, B

Samali, P Abtahi, 6th WCSCM: “World Conference on Structural Control and

Monitoring”, Barcelona Spain, July 2014.

4. Evaluation of In-plane and out-of-plane movement of façade panels to reduce

structure response during earthquake excitation, P Abtahi, B Samali, 23th

ACMSM: “Materials to Structures: Advancement through Innovation”, Byron Bay,

Australia December 2014.

5. A Review of the Drawbacks of Current Fixed Connection Façade Systems, Non-

Structural Standards, and Ways of Integrating Movable Façade Technology into

Buildings, P Abtahi, B Samali, ICACE 2015: “International Conference on

Architecture and Civil Engineering” Venice, Italy, April 2015.

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TABLE OF CONTENTS

CERTIFICATE OF AUTHORSHIP/ORIGINALITY ....................................................................................... I

ACKNOWLEDGMENTS ............................................................................................................................... II

CHAPTER 1 ..................................................................................................................................................... 1

INTRODUCTION

1.1 BACKGROUND OF THE STUDY .......................................................................................................... 2

1.2 RESEARCH PROBLEM ....................................................................................................................... 5

1.3 RESEARCH OBJECTIVE AND AIMS .................................................................................................... 6

1.3.1 Research Objectives ................................................................................................................... 6

1.3.2 Thesis Aims ............................................................................................................................... 8

1.4 METHODOLOGY OF RESEARCH ........................................................................................................ 8

1.5 SCOPE OF RESEARCH ....................................................................................................................... 9

1.6 DISSERTATION LAYOUT .................................................................................................................. 9

CHAPTER 2 ................................................................................................................................................... 11

GENERAL INFORMATION ABOUT EARTHQUAKE LOADS AND METHODS OF

MITIGATING SEISMIC ACTIVITY

2.1 INTRODUCTION .............................................................................................................................. 12

2.2 EARTHQUAKE-RESISTANT DESIGN FOR STRUCTURAL BUILDINGS ................................................... 13

2.2.1 Force Method ........................................................................................................................... 17

2.2.2 Displacement Method .............................................................................................................. 17

2.2.3 Review of AS1170.4 Australian Standard ............................................................................... 18

2.2.4 Seismic retrofitting of building structure ................................................................................. 18

2.3 DRAWBACKS OF CURRENT NON-STRUCTURAL AUSTRALIAN AND INTERNATIONAL STANDARDS

FOR THE DESIGN AND DAMAGE ASSESSMENT OF PRIMARY STRUCTURE AND FAÇADE PANELS ....................... 19

2.3.1 Introduction .............................................................................................................................. 19

2.4 MODIFICATIONS IN STRUCTURAL SYSTEMS ................................................................................... 29

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2.4.1 Cladding Isolation .................................................................................................................... 30

2.4.2 Addition of Damping Systems ................................................................................................. 30

2.5 ENERGY DISSIPATION SYSTEMS .................................................................................................... 32

2.5.1 Passive Controllers ................................................................................................................... 33

2.5.1.1 Tuned Mass Damper ....................................................................................................... 34

2.5.1.1.1 Equation of Motion ..................................................................................................... 36

2.5.1.1.2 Determining TMD Parameters .................................................................................... 40

2.5.1.2 Tuned Liquid dampers .................................................................................................... 40

2.5.1.3 Multiple Tuned Mass Dampers ...................................................................................... 42

2.5.1.4 Nonlinear Tuned Mass Dampers (NTMD) ..................................................................... 46

2.5.1.5 Pendulum Tuned Mass Damper (PTMD) ....................................................................... 49

2.5.1.6 Base Isolation ................................................................................................................. 51

2.5.1.7 Viscous Fluid Dampers (VF) .......................................................................................... 52

2.5.1.8 Viscoelastic Dampers (VE) ............................................................................................ 53

2.5.2 Semi-Active Controllers .......................................................................................................... 54

2.5.3 Active Control of Structures .................................................................................................... 55

2.6 ANALYTICAL METHOD FOR ANALYSING NONLINEAR SYSTEMS .................................................... 57

2.6.1 Perturbation Method: Multiple Scales Method ........................................................................ 58

2.6.2 Local Stability Analysis ........................................................................................................... 59

2.7 NUMERICAL METHODS FOR ANALYSING NONLINEAR SYSTEMS .................................................... 60

2.7.1 Time Integration Method ......................................................................................................... 61

2.7.2 Continuation Method ............................................................................................................... 62

2.8 SUMMARY ..................................................................................................................................... 65

CHAPTER 3 ................................................................................................................................................... 67

LITERATURE REVIEW OF FAÇADE SYSTEMS

3.1 INTRODUCTION .............................................................................................................................. 68

3.2 TYPES OF FACADE SYSTEMS .......................................................................................................... 70

3.2.1 Infills ........................................................................................................................................ 70

3.2.2 Light Weight Cladding ............................................................................................................. 71

3.2.2.1 Stick System ................................................................................................................... 71

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3.2.2.2 Curtain Wall ................................................................................................................... 72

3.2.2.3 Unitized Curtain Wall ..................................................................................................... 73

3.2.2.4 Panelized Curtain Wall ................................................................................................... 75

3.2.2.5 Spandrel Panel Ribbon Glazing ...................................................................................... 75

3.2.2.6 Bolted Glass Façade ....................................................................................................... 76

3.2.2.7 Double Skin Façade ........................................................................................................ 77

3.2.2.7.1 Definition .................................................................................................................... 77

3.2.2.7.2 History ........................................................................................................................ 80

3.2.2.7.3 Examples ..................................................................................................................... 81

3.2.3 Heavyweight Cladding ............................................................................................................. 82

3.3 TYPICAL FACADE CONNECTIONS AND THEIR INHERENT PROBLEMS .............................................. 83

3.3.1 Bearing Connection .................................................................................................................. 84

3.3.2 Tie-back connection ................................................................................................................. 84

3.3.3 Governing Failure Mechanism in Attachment ......................................................................... 84

3.4 FAÇADE PANELS CAPABILITY AND COMPATIBILITY TO THE PROPOSED NOVEL DESIGNS.............. 87

3.4.1 Infills ........................................................................................................................................ 88

3.4.2 Lightweight Cladding .............................................................................................................. 89

3.4.3 Curtain Walls ........................................................................................................................... 90

3.4.4 Panelised Curtain Wall ............................................................................................................. 91

3.4.5 Double Skin Façade System..................................................................................................... 92

3.4.6 Heavyweight Cladding ............................................................................................................. 92

3.5 CHAPTER SUMMERY AND CONCLUSION ........................................................................................ 94

CHAPTER 4 ................................................................................................................................................... 96

FEASIBILITY STUDY & PRIMARY NUMBERICAL MODELLING

4.1 INTRODUCTION AND METHODOLOGY ............................................................................................ 97

4.2 BEHAVIOUR OF DOUBLE-SKIN-FAÇADE IN SUPPRESSING EARTHQUAKE LOADS ........................... 99

4.2.1 Introduction .............................................................................................................................. 99

4.2.2 System Modelling .................................................................................................................. 100

4.2.3 Dynamic Responses of the System ........................................................................................ 101

vii

4.3 FIRST PROPOSAL (FEASIBILITY STUDY OF FAÇADE SYSTEM AS MULTI TUNED MASS DAMPER WITH

3D NUMERICAL MODELLING IN SAP2000) .................................................................................................... 103

4.3.1 Earthquake Loading Records and Boundary Condition: ........................................................ 104

4.3.2 Material Properties ................................................................................................................. 108

4.3.3 Structural Modelling and its Dynamic Behaviour .................................................................. 109

4.3.4 Results of Computer Modelling ............................................................................................. 112

4.3.5 Conclusion: ............................................................................................................................ 114

4.4 SECOND PROPOSAL - NUMERICAL MODELLING OF FACADES WITH SACRIFICIAL ELEMENTS IN

SAP2000 115

4.4.1 Introduction ............................................................................................................................ 115

4.4.2 Preliminary Numerical Modelling ......................................................................................... 116

4.4.3 Out-of-Plane Concept of Façade Behaviour .......................................................................... 119

4.4.3.1 Results and Discussion ................................................................................................. 121

4.4.3.2 Conclusion .................................................................................................................... 122

4.4.4 In-Plane Concept of Façade Behaviour .................................................................................. 122

4.4.4.1 Results and Discussion ................................................................................................. 125

4.4.4.2 Conclusion .................................................................................................................... 127

4.5 FINITE ELEMENT MODELLING USING ANSYS APDL ................................................................. 128

4.5.1 Introduction of Smart Bracket with Combined Shear and Axial Movement ......................... 128

4.5.2 Development of Smart Passive Façade System-Assigning a Nonlinear Behaviour to Façade

Connection 129

4.5.3 Earthquake Records and Their Features ................................................................................. 131

4.5.4 Structural Models ................................................................................................................... 133

4.5.5 Bracket Element Behaviour ................................................................................................... 133

4.5.6 Criteria for Evaluation of the System ..................................................................................... 135

4.5.6.1 Lateral Displacement Control ....................................................................................... 135

4.5.6.2 Drift Control ................................................................................................................. 138

4.5.6.3 Acceleration of Primary Structure ................................................................................ 141

4.5.6.4 Root Mean Square of Top Displacement ...................................................................... 142

4.5.7 Conclusions ............................................................................................................................ 143

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CHAPTER 5 ................................................................................................................................................. 145

MID-RISE 10-STOREY STRUCTRAL MODELS

INTRODUCTION

5.2 MODELLING APPROACH AND ASSUMPTIONS ............................................................................... 146

5.2.1 Finite Element Analysis ......................................................................................................... 147

5.2.1.1 Meshing Size ................................................................................................................ 147

5.2.1.1.1 Column, Beam and Slab ............................................................................................ 148

5.2.1.1.2 Bracket (Spring Beam) .............................................................................................. 148

5.2.1.1.3 Façade Column ......................................................................................................... 148

5.2.1.2 Direction of Applied Loads and Building Boundary Conditions ................................. 148

5.2.2 Element Behaviour and Structural Modelling ........................................................................ 149

5.2.2.1 Material Properties ....................................................................................................... 151

5.2.2.2 Consideration of Seismic Mass .................................................................................... 151

5.2.2.3 Design of Structural Elements ...................................................................................... 152

5.2.2.3.1 Equivalent Sections for Structural Beams and Columns ........................................... 152

5.2.2.4 Assumption of Strong Column and Weak Beam Connection by Semi-Rigid Connections

153

5.2.3 Details of Cladding System (Double Skin) Modelling........................................................... 154

5.2.3.1 Material Properties ....................................................................................................... 155

5.2.3.2 Façade Column Modelling ........................................................................................... 155

5.2.3.3 Details of Bracket Modelling ....................................................................................... 158

5.3 TYPES OF NUMERICAL MODELLINGS ........................................................................................... 161

5.3.1 Modal Analysis ...................................................................................................................... 161

5.3.1.1 Results .......................................................................................................................... 161

5.3.2 Dynamic Time-History Analysis (Holistic Nonlinear Time-History Analysis) ..................... 163

5.3.2.1 Dynamic Interpretation of Energy Content the Selected Earthquakes (Power Spectral

Density) 163

5.3.2.2 Judgment on Engineering Demand Parameters ............................................................ 167

5.3.2.2.1 Top Lateral Displacement ......................................................................................... 168

5.3.2.2.2 Root Mean Square (RMS) of Top Displacement (mm) ............................................ 176

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5.3.2.2.3 Relative Displacement of Façade and Main Structure .............................................. 177

5.3.2.2.4 Structural Inter-Storey Drift ...................................................................................... 179

5.3.2.2.4.1 Inter-Storey Drift Calculation and In-Plane Seismic Design ............................. 179

5.3.2.2.4.2 Results ............................................................................................................... 180

5.3.2.2.5 Top Lateral Acceleration........................................................................................... 184

5.3.2.2.6 Root Mean Square (RMS) of Top Acceleration ........................................................ 191

5.3.2.2.7 Base Shear ................................................................................................................. 192

5.4 FINDINGS AND CONCLUSION ....................................................................................................... 193

CHAPTER 6 ................................................................................................................................................. 194

HIGH-RISE 30-STOREY STRUCTRAL MODELS

6.1 INTRODUCTION ............................................................................................................................ 195

6.2 ASSUMPTIONS.............................................................................................................................. 196

6.2.1 Element behaviour and structural modelling ......................................................................... 196

6.2.2 Details of Bracket modelling ................................................................................................. 198

6.3 TYPES OF NUMERICAL MODELLINGS ............................................................................................ 199

6.3.1 Modal Analysis ...................................................................................................................... 199

6.3.1.1 Results .......................................................................................................................... 200

6.3.2 Dynamic Time-History Analysis (Holistic Nonlinear time-history analysis) ........................ 201

6.3.2.1 Judgement on engineering demand parameters ............................................................ 202

6.3.2.1.1 Top lateral displacement ........................................................................................... 202

6.3.2.1.2 Root mean square (RMS) response of top displacement (mm) ................................. 210

6.3.2.1.3 Maximum relative displacement of façade and main structure ................................. 211

6.3.2.1.4 Structural inter-storey drift ........................................................................................ 212

6.3.2.1.5 Top lateral acceleration ............................................................................................. 217

6.3.2.1.6 Root mean square (RMS) of top acceleration response ............................................ 223

6.3.2.1.7 Base shear force ........................................................................................................ 224

6.4 APPLICATION OF ADVANCED CLADDING CONNECTIONS AND DESIGN STEPS .............................. 225

6.5 FINDINGS AND CONCLUSION ........................................................................................................ 226

CHAPTER 7 ................................................................................................................................................. 228

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FINANCIAL CONSIDERATIONS

7.1 INTRODUCTION ............................................................................................................................ 229

7.2 FINANCIAL ASSESSMENT OF MOVABLE FAÇADE SYSTEM AS A NEW RETROFITTING METHOD ....... 230

7.3 ADDITIONAL COST OF THE MOVABLE FACADE TO BUILDING STRUCTURE ..................................... 230

7.3.1 Introduction ............................................................................................................................ 230

7.3.2 Design or re-design procedure ............................................................................................... 231

7.3.3 Maintenance ........................................................................................................................... 232

7.3.3.1 Preventive maintenance strategies and their cost.......................................................... 232

7.3.3.2 Quarterly and annual inspection of the proposed system ............................................. 234

7.3.4 Importance of thermal performance ....................................................................................... 236

7.3.5 Damage cost to the façade system after an earthquake .......................................................... 236

7.3.6 Interaction with insurance companies for Earthquake Insurance Premium (EIP) .................. 237

7.4 COST BENEFITS TO THE MAIN STRUCTURE ................................................................................... 238

7.4.1 Introduction ............................................................................................................................ 238

7.4.2 Case studies ............................................................................................................................ 244

7.4.2.1 Case study 1 .................................................................................................................. 244

7.4.2.1.1 Construction cost....................................................................................................... 245

7.4.2.1.2 Construction time ...................................................................................................... 246

7.4.2.1.3 Labour cost ................................................................................................................ 249

7.4.2.1.4 Rental income ........................................................................................................... 249

7.4.2.1.5 Overall Saving .......................................................................................................... 250

7.4.2.2 Case study 2 .................................................................................................................. 252

7.4.2.2.1 Construction cost....................................................................................................... 252

7.4.2.2.2 Construction time ...................................................................................................... 254

7.4.2.2.3 Labour cost ................................................................................................................ 254

7.4.2.2.4 Rental income ........................................................................................................... 255

7.4.2.2.5 Overall Profit ............................................................................................................ 256

7.5 STRATEGIES AND APPROACHES .................................................................................................... 258

7.6 SUMMARY AND CONCLUSIONS ..................................................................................................... 258

CHAPTER 8 ................................................................................................................................................. 263

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CONCLUSION AND FUTURE WORK

8.1 GENERAL CONCLUSIONS .............................................................................................................. 264

8.1.1 Application and contribution of this research to design ......................................................... 268

8.2 RECOMMENDATIONS FOR FUTURE RESEARCH .............................................................................. 270

8.2.1 Further research that would improve and complement this thesis ......................................... 271

8.2.2 Proposed experimental test program ...................................................................................... 272

8.2.2.1 Test setup, specimen design and terminology .............................................................. 272

APPENDIX A ............................................................................................................................................... 279

SECTIONS OF THE STRUCTURAL MODELS ........................................................................................ 279

APPENDIX B ............................................................................................................................................... 288

THESIS TERMINOLOGY ........................................................................................................................... 288

REFERENCE ................................................................................................................................................ 291

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LIST OF FIGURES

Figure 2-1: Seismic hazard versus return period (Paulay 1992) ........................................................................... 12

Figure 2-2: Seismic performance levels of a building .......................................................................................... 25

Figure 2-3: Relationship between excitation and role of passive controllers in structure Loop (Symans 1999) .. 34

Figure 2-4: Schematic View of Displacement of TMD (Soong and Spencer 2002) ............................................. 36

Figure 2-5: Single Degree of Freedom System with the Use of TMD (Soong and Dargush 1997)...................... 37

Figure 2-6: Multi-Degree-of-Freedom System with TMD ................................................................................... 39

Figure 2-7 Illustration of a schematic model of a TLD ........................................................................................ 41

Figure 2-8: TLD used in Rincon Hill (the first U.S. residential tower) (Soong and Spencer 2002) ..................... 42

Figure 2-9 Schematic model of multiple TMD (MTMDs) in parallel .................................................................. 42

Figure 2-10 Schematic model of multiple TMD (MTMDs) in series ................................................................... 43

Figure 2-11 Schematic model of nonlinear TMD (NTMD) ................................................................................. 46

Figure 2-12 Illustration of the PTMD installed in Taipei 101 (Soong and Spencer 2002) ................................... 49

Figure 2-13: Difference between lateral deformation in controlled and uncontrolled systems (Otani 1981) ....... 51

Figure 2-14: Typical viscous fluid dampers used in diagonal bracings (Hemalatha and Jaya 2008) ................... 53

Figure 2-15: Typical VE Damper Configuration (Soong and Spencer 2002) ....................................................... 54

Figure 2-16: Structure with a Semi-Active Control System (Symans 1999) ........................................................ 55

Figure 2-17: Structure with Active Control System (Symans 1999) .................................................................... 56

Figure 2-18: Structure with a Hybrid Control System (Symans 1999) ................................................................. 57

Figure 3-1: Typical components of a façade panel (Olanders Window Replacement 2011) ................................ 68

Figure 3-2: Stick system façade (Permasteelisa 2009) ......................................................................................... 71

Figure 3-3: Typical assembly of stick system façade (Permasteelisa 2009) ......................................................... 72

Figure 3-4: Unitized Curtain Wall (Permasteelisa 2009) ...................................................................................... 74

Figure 3-5: Installation of curtain wall (Permasteelisa 2009) ............................................................................... 74

Figure 3-6: Panelized curtain wall (Permasteelisa 2009) ...................................................................................... 75

Figure 3-7: Example of spandrel panel ribbon glazing (Permasteelisa 2009) ...................................................... 76

Figure 3-8: Independent assembly (Permasteelisa 2009) ..................................................................................... 76

Figure 3-9: Suspended assembly (Permasteelisa 2009) ........................................................................................ 77

Figure 3-10: Typical Double-Skin-Façade System (Poizaris 2004) ..................................................................... 78

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Figure 3-11: Exterior Circulation Double Skin Curtain Wall (Arons 2000) ........................................................ 79

Figure 3-12: Facade detail: Hot expelled at each floor, cool air drawn in (Lee, Selkowitz et al. 2002) ............... 79

Figure 3-13: Steiff factory, Giengen/Brenz, Germany (Streicher, Heimrath et al. 2007). .................................... 81

Figure 3-14: Precast facade panel installations (Traulsen and McClellan 2010) .................................................. 82

Figure 3-15: Different failure mechanisms and push-over behaviour of precast panels attached to a frame system

(Baird, Diaferia et al. 2011) ........................................................................................................................ 85

Figure 4-1: Simplified model of the primary structure and façade system connected by movable brackets ...... 100

Figure 4-2: Detail of façade connection to primary structure and modelling assumption in SAP2000 .............. 101

Figure 4-3: Scaled Northridge earthquake excitation record .............................................................................. 105

Figure 4-4: Scaled El-Centro earthquake excitation record ................................................................................ 105

Figure 4-5: Scaled Kobe earthquake excitation record ....................................................................................... 105

Figure 4-6: Scaled Hachinohe earthquake excitation record .............................................................................. 106

Figure 4-7: Response Spectra of scaled Northridge earthquake record .............................................................. 106

Figure 4-8: Response Spectra of scaled El Centro earthquake record ................................................................ 107

Figure 4-9: Response Spectra of scaled Kobe earthquake record ....................................................................... 107

Figure 4-10: Response Spectra of scaled Hachinohe earthquake record ............................................................ 107

Figure 4-11: Schematic plan of the structural model .......................................................................................... 109

Figure 4-12: Schematic view of primary structural model with details of brackets and connections ................. 111

Figure 4-13: Time history analysis of structural model under Northridge earthquake ....................................... 112

Figure 4-14: Time history analysis of structural model under El-Centro earthquake ......................................... 112

Figure 4-15: Time history analysis of structural model under Kobe earthquake ................................................ 113

Figure 4-16: Time history analysis of structural model under Hachinohe earthquake ....................................... 113

Figure 4-17: Elastic structure in X direction ....................................................................................................... 117

Figure 4-18: plastic structure with auto-defined plastic hinges in elements ....................................................... 118

Figure 4-19: plastic structure incorporated with façade elements with auto-defined plastic hinges in structural

elements and user-defined plastic hinges in bracket elements .................................................................. 118

Figure 4-20: Typical façade panel subjected to wind forces............................................................................... 120

Figure 4-21: Top lateral displacement of structure with plastic brackets in El-Centro earthquake .................... 121

Figure 4-22: Rheinbach glass museum, Rheinbach (Wellershoff and Sedlacek 2003) ...................................... 123

xiv

Figure 4-23: In-plane concept of façade behaviour (façade as a shell element in structure frame and its

connections) .............................................................................................................................................. 124

Figure 4-24: Relative displacement of top level of main structure with different stiffness of bracket under 1940

El Centro Earthquake ................................................................................................................................ 125

Figure 4-25: Relative displacement between shell and structure at top level under 1940 El Centro Earthquake126

Figure 4-26: Proposed Multi-linear behaviour of the façade bracket acting as axial damper system ................. 129

Figure 4-27: Seismic hazard versus return period (Paulay 1992) ....................................................................... 131

Figure 4-28: Displacement Power Spectrum Density for 1994 Northridge earthquake ..................................... 132

Figure 4-29: Displacement Power Spectrum Density for 1963 Hachinohe earthquake ...................................... 132

Figure 4-30: Configuration of the proposed damper system .............................................................................. 134

Figure 4-31: Relative displacement between top and bottom of the primary structure during Northridge record

.................................................................................................................................................................. 136

Figure 4-32: Relative displacement between top and bottom of the primary structure during Hachinohe record

.................................................................................................................................................................. 137

Figure 4-33: Drift for primary structure with different stiffness for shear bracket façade elements during

Northridge earthquake ............................................................................................................................... 140

Figure 4-34: Drift for primary structure with different stiffness for shear bracket façade elements during

Hachinohe earthquake ............................................................................................................................... 141

Figure 4-35: Acceleration in top floor of primary structure with different shear stiffness for bracket facades

during Northridge earthquake ................................................................................................................... 142

Figure 4-36: Acceleration in top floor of primary structure with different shear stiffness for bracket facades

during Hachinohe earthquake ................................................................................................................... 142

Figure 5-1: Direction of applied earthquake in the 3-D model ........................................................................... 149

Figure 5-2: Front view of exterior elevation of the 3D frame model .................................................................. 150

Figure 5-3: Plan view of the 3D frame model (dimensions are in mm).............................................................. 150

Figure 5-4: Reinforced concrete section and assumed equivalent section .......................................................... 153

Figure 5-5: Details of proposed column/beam connection ................................................................................. 154

Figure 5-6: Modelled force-deformation hysteretic curve for modelling plastic hinges in reinforced concrete

beams ........................................................................................................................................................ 154

Figure 5-7: Schematic view of facade column element and their configuration in each floor ........................... 157

xv

Figure 5-8: Elevation view of façade connection ............................................................................................... 158

Figure 5-9: Defined Force-deformation curve of the axial connection ............................................................... 159

Figure 5-10: Defined Force-deformation curve of the shear connection ............................................................ 160

Figure 5-11: Plan view of damper connections to the main structure and their behaviour in applied earthquake

.................................................................................................................................................................. 160

Figure 5-12: Comparison of effective modal mass percentages for the first three modes .................................. 162

Figure 5-13: Seismic and wind hazard versus excitation frequency (or period) (Paulay 1992) ......................... 164

Figure 5-14: Displacement Power Spectral Density for 1994 Northridge earthquake ........................................ 165

Figure 5-15: Displacement Power Spectrum Density for 1940 El-Centro earthquake ....................................... 166

Figure 5-16: Displacement Power Spectrum Density for 1995 Kobe earthquake .............................................. 167

Figure 5-17: Displacement Power Spectrum Density for Hachinohe earthquake............................................... 167

Figure 5-18: Time-history of top floor displacements of primary structure coupled with DSFs with different

bracket connector stiffness during 1994 Northridge earthquake ............................................................... 169

Figure 5-19: Top floor displacements of primary structure coupled with DSFs with different bracket connector

stiffness during 1940 El-Centro Earthquake ............................................................................................. 170


stiffness during 1995 Kobe Earthquake .................................................................................................... 170


stiffness during 1968 Hachinohe Earthquake ............................................................................................ 171

Figure 5-22: Top floor displacements of primary structure coupled with DSFs with different shear connector

during 1994 Northridge Earthquake .......................................................................................................... 172


during 1940 El-Centro Earthquake ........................................................................................................... 173


during 1995 Kobe Earthquake .................................................................................................................. 174


during 1968 Hachinohe Earthquake .......................................................................................................... 175

Figure 5-26: Schematic diagram of a building movement under earthquake ground motion. ............................ 180

Figure 5-27: Drift for primary structure with different stiffness for shear bracket façade elements during 1994


xvi


El-Centro earthquake ................................................................................................................................ 181


Kobe earthquake ....................................................................................................................................... 182



Figure 5-31: Time-history of top floor accelerations of primary structure coupled with DSFs with optimal

bracket connector stiffness during 1994 Northridge Earthquake .............................................................. 185

Figure 5-32: Time-history top floor accelerations of primary structure coupled with DSFs with optimal bracket

connector stiffness during 1940 El-Centro Earthquake ............................................................................. 186

Figure 5-33 : Time-history top floor accelerations of primary structure coupled with DSFs with optimal bracket

connector stiffness during 1995 Kobe Earthquake .................................................................................... 186


connector stiffness during Hachinohe Earthquake .................................................................................... 187

Figure 5-35: Calculation of base shear in a structure.......................................................................................... 192

Figure 6-1: Front view of exterior elevation of the 3D frame model .................................................................. 197

Figure 6-2: Plan view of the 30m by 30m 3D frame model (dimensions are in mm) ........................................ 197

Figure 6-3: Defined Force-deformation curve of the axial connection ............................................................... 198

Figure 6-4: Defined Force-deformation curve of the shear connection .............................................................. 199

Figure 6-5: Comparison of effective modal mass percentages for the first three modes .................................... 201


bracket connector stiffness during 1994 Northridge earthquake ............................................................... 203


bracket connector stiffness during 1940 El Centro earthquake ................................................................. 204


stiffness during 1995 Kobe Earthquake .................................................................................................... 204


stiffness during 1968 Hachinohe Earthquake ............................................................................................ 205

Figure 6-10: Top floor displacements of primary structure coupled with DSFs with different shear connectors

during 1994 Northridge Earthquake .......................................................................................................... 206

xvii


during 1940 El Centro Earthquake ............................................................................................................ 207


during 1995 Kobe Earthquake .................................................................................................................. 208


during 1968 Hachinohe Earthquake .......................................................................................................... 208

Figure 6-14: Maximum drift for primary structure with different stiffness for shear bracket façade elements

during 1994 Northridge earthquake .......................................................................................................... 213


during 1940 El-Centro earthquake ............................................................................................................ 214


during 1995 Kobe earthquake ................................................................................................................... 214


during 1968 Hachinohe earthquake ........................................................................................................... 215

Figure 6-18: Time-history of top floor accelerations of primary structure coupled with DSFs with optimal

bracket connector stiffness during 1994 Northridge Earthquake .............................................................. 218


connector stiffness during 1940 El-Centro Earthquake ............................................................................. 218


connector stiffness during 1995 Kobe Earthquake .................................................................................... 219


connector stiffness during Hachinohe Earthquake .................................................................................... 219

Figure 7-1: Earthquake Insurance Premium versus seismic zone (Permasteelisa 2009) .................................... 238

Figure 7-2: Location of selected cities ................................................................................................................ 240

Figure 7-3: Construction costs ............................................................................................................................ 242

Figure 7-4: Labour costs ..................................................................................................................................... 242

Figure 7-5: Material costs ................................................................................................................................... 243

Figure 7-6: Rental prices and cap rate ................................................................................................................ 243

Figure 7-7: Comparison between dimensions of beam/column elements in both the conventional and smart

façade systems........................................................................................................................................... 244

xviii

Figure 7-8: Building Construction cost with conventional façade and smart façade systems ............................ 245

Figure 7-9: Construction cost distribution with conventional and smart façade systems ................................... 246

Figure 7-10: Comparison of construction time between conventional and smart façade systems ...................... 248

Figure 7-11: Construction time with constant amount of labour and time in both smart and conventional façade

systems. ..................................................................................................................................................... 248

Figure 7-12: Comparison of labour costs between conventional and smart façade systems .............................. 249

Figure 7-13: Rental income increase due to smart façade system over 20 years ................................................ 250

Figure 7-14: Comparison of building component expenses profit by using smart façade system ...................... 251

Figure 7-15: Comparison of overall saving by using smart façade system ........................................................ 251

Figure 7-16: Comparison between dimensions of beam/column elements in both the conventional and smart

façade systems........................................................................................................................................... 252

Figure 7-17: Building Construction cost with conventional façade and smart façade systems .......................... 253

Figure 7-18: Construction cost distribution of a mid-rise building with conventional and smart façade systems

.................................................................................................................................................................. 253

Figure 7-19: Comparison of construction time between conventional and smart façade systems ...................... 254

Figure 7-20: Comparison of labour cost between conventional and smart façade systems ............................... 255

Figure 7-21: Additional rental income due to using smart façade system .......................................................... 256

Figure 7-22: Comparison of building component expenses/savings by using smart façade system ................... 257

Figure 7-23: Comparison of overall savings when using smart façade system .................................................. 257

Figure 8-1: Details of proposed connection for attachment of façade outer skin to slab of main structure ........ 271

Figure 8-2: South-west sketch of the building structure and elevation of the specimen ..................................... 273

Figure 8-3: Sketch of details of experimental model .......................................................................................... 274





Figure 8-8: Side views of attachment of the proposed damper system to slab of primary structure .................. 277

Figure 8-9: Top view of attachment of damper system to slab of primary structure .......................................... 278

xix

LIST OF TABLES

Table 2-1: Non-structural performance level ........................................................................................................ 25

Table 3-1: Different kinds of façade systems ....................................................................................................... 70

Table 4-1: Earthquake ground motions used in this study .................................................................................. 105

Table 4-2: Material properties of façade system ................................................................................................. 108

Table 4-3: Material properties of primary structure............................................................................................ 108

Table 4-4: Structure model dynamic properties .................................................................................................. 110

Table 4-5: Results of time history analysis ......................................................................................................... 113

Table 4-6: Frequency of first three modes of the structure ................................................................................. 116

Table 4-7: Root mean square of top floor displacement cases with different plastic plateau forces (mm) ........ 122

Table 4-8: Maximum lateral displacement of the structure in different value of shell spring stiffness in El Centro

Earthquake ................................................................................................................................................ 126

Table 4-9: Root Means Square for in-plane-shell ............................................................................................... 126

Table 4-10: Lateral Displacement of top façade panels under 1940 El Centro Earthquake ............................... 127

Table 4-11: Root Mean Square of relative displacement of façade shell element at top level of structure ........ 127

Table 4-12: Maximum top lateral displacement in primary structure incorporating low shear stiffness bracket

facades during Northridge record .............................................................................................................. 137

Table 4-13: Maximum top lateral displacement in primary structure incorporating with low shear stiffness

bracket facades during Hachinohe record ................................................................................................. 138

Table 4-14: In plane drift of primary structure with different bracket stiffness during 1994 Northridge

earthquake (in mm) ................................................................................................................................... 140

Table 4-15: In plane drift of primary structure in case of different bracket stiffness during 1963 Hachinohe

earthquake (in mm) ................................................................................................................................... 141

Table 4-16: Root mean square of top displacement using different value of shear stiffness during 1994


Table 4-17: Root mean square of top displacement using different value of shear stiffness during 1963


Table 5-1: Structural sections for beam and column elements ........................................................................... 149

Table 5-2: Selected concrete properties .............................................................................................................. 151

xx

Table 5-3: Mass values for bare frame model .................................................................................................... 151

Table 5-4: Assumed factors for earthquake design ............................................................................................. 152

Table 5-5: Material properties of façade panel components ............................................................................... 155

Table 5-6: Modal vibration periods of models with bare-frame, fixed and flexible facades .............................. 161

Table 5-7: Participating modal mass percentages ............................................................................................... 162

Table 5-8: Characteristics of selected earthquake records .................................................................................. 164

Table 5-9: Comparison between maximum top floor displacements of primary structure coupled with DSFs with

different bracket connector stiffness during the four earthquakes............................................................. 171

Table 5-10: Comparison between maximum top floor displacements of primary structure with various shear

façade bracket stiffness during the four earthquakes ................................................................................ 176

Table 5-11: Root mean square of top displacement using different value of shear stiffness for the 3D models

during the excitations ................................................................................................................................ 177

Table 5-12: Relative Displacement between the primary structure model and outer layer of façade system during

1994 Northridge earthquake ...................................................................................................................... 177


1940 El-Centro earthquake ....................................................................................................................... 178


1995 Kobe earthquake .............................................................................................................................. 178

Table 5-15: Relative Displacement between the primary structure model and out layer of façade system during

1968 Hachinohe earthquake ...................................................................................................................... 178

Table 5-16: Comparison of storey drift with different bracket stiffness during 1994 Northridge earthquake .... 183

Table 5-17: Comparison of storey drift with different bracket stiffness during 1940 El Centro earthquake ...... 183

Table 5-18: Comparison of storey drift with different bracket stiffness during 1995 Kobe earthquake ............ 183

Table 5-19: Comparison of storey drift with different bracket stiffness during 1968 Hachinohe earthquake .... 184

Table 5-20: Comparison of absolute maximum value of interstorey drifts for each bracket case ...................... 184

Table 5-21: Comparison between top floor accelerations (mm/sec2) of primary structure coupled with DSFs with

optimal bracket connector stiffness during 1994 Northridge Earthquake ................................................. 187


optimal bracket connector stiffness during 1940 El Centro Earthquake ................................................... 188

xxi


optimal bracket connector stiffness during 1995 Kobe Earthquake .......................................................... 188


optimal bracket connector stiffness during 1968 Hachinohe Earthquake ................................................. 189

Table 5-25: Comparison between top floor accelerations (mm/sec2) of primary structure with various shear

façade bracket stiffness during 1994 Northridge earthquake .................................................................... 189


façade bracket stiffness during 1940 El Centro earthquake ...................................................................... 190


façade bracket stiffness during 1995 Kobe earthquake ............................................................................. 190


façade bracket stiffness during 1968 Hachinohe earthquake .................................................................... 190

Table 5-29: Root mean square of top acceleration using optimal values of shear stiffness for the 3D models

during the four excitations ........................................................................................................................ 191

Table 5-30: Comparison between base shear forces (Knaack, Klein et al.) of the 3-D primary structure in various

bracket stiffness......................................................................................................................................... 193

Table 6-1: Structural sections for beam and column elements ........................................................................... 196

Table 6-2: Modal vibration periods of models with bare-frame, fixed and flexible facades .............................. 200

Table 6-3: Participating modal mass percentages ............................................................................................... 201

Table 6-4: Comparison between maximum top floor displacements of primary structure coupled with DSFs with

different bracket connector stiffness during the four earthquakes............................................................. 205

Table 6-5: Comparison between maximum top floor displacements (mm) of primary structure with various shear

façade bracket stiffness during the four earthquakes ................................................................................ 209

Table 6-6: Root mean square response of top displacement using different values of shear stiffness for the 3D

models during different earthquake excitations ........................................................................................ 210

Table 6-7: Maximum relative Displacement between 3D structure model and outer layer of façade system

during 1994 Northridge earthquake .......................................................................................................... 211


during 1940 El-Centro earthquake ............................................................................................................ 211

xxii


during 1995 Kobe earthquake ................................................................................................................... 212

Table 6-10: Maximum relative Displacement between 3D structure model and out layer of façade system during

1968 Hachinohe earthquake ...................................................................................................................... 212

Table 6-11: Comparison of maximum inter-storey drift with different bracket stiffness during 1994 Northridge

earthquake ................................................................................................................................................. 215

Table 6-12: Comparison of maximum inter-storey drift with different bracket stiffness during 1940 El-Centro

earthquake ................................................................................................................................................. 216

Table 6-13: Comparison of maximum inter-storey drift with different bracket stiffness during 1995 Kobe

earthquake ................................................................................................................................................. 216

Table 6-14: Comparison of maximum inter storey drift with different bracket stiffness during 1968 Hachinohe

earthquake ................................................................................................................................................. 216

Table 6-15: Comparison of absolute maximum values of inter-storey drifts for each bracket case ................... 217


optimal bracket connector stiffness during 1994 Northridge Earthquake ................................................. 220


optimal bracket connector stiffness during 1940 El Centro Earthquake ................................................... 220


optimal bracket connector stiffness during 1995 Kobe Earthquake .......................................................... 221


optimal bracket connector stiffness during 1968 Hachinohe Earthquake ................................................. 221

Table 6-20: Comparison between top floor accelerations of primary structure with various shear façade bracket

stiffness during 1994 Northridge earthquake ............................................................................................ 222


stiffness during 1940 El Centro earthquake .............................................................................................. 222


stiffness during 1995 Kobe earthquake ..................................................................................................... 222


stiffness during 1968 Hachinohe earthquake ............................................................................................ 223

xxiii

Table 6-24: Root mean square of top acceleration (mm/sec2) using optimal values of shear stiffness for the 3D

models during the four excitations ............................................................................................................ 224

Table 6-25: Comparison between base shear (Knaack, Klein et al.) of the primary structure with various bracket

shear stiffness ............................................................................................................................................ 225

Table 7-1: Existing methods of retrofitting ........................................................................................................ 229

Table 7-2: Details of additional price of smart façade system ............................................................................ 232

Table 7-3: Proposed quarterly and yearly spreadsheet for inspection of each damper/connector component .... 235

Table 7-4: Spreadsheet for expected yearly expenses per square meter of façade panel .................................... 235

Table 7-5: Damage cost and damage state of the movable façade system located on top level of a mid-rise

structure ..................................................................................................................................................... 237

Table 7-6: Three major approaches in measuring floor area .............................................................................. 241

Table 7-7: Rent definitions ................................................................................................................................. 241

Table 7-8: General specifications of structural models ...................................................................................... 244

Table 7-9: Comparison of economic impacts between conventional and smart façade systems ........................ 246

Table 7-10 : Investigated parameter – Construction time ................................................................................... 247

Table 7-11: Investigated parameter – Construction time .................................................................................... 247

Table 7-12: Values (million$) of building component savings by using the smart façade system ..................... 251

Table 7-13: Comparison of economic benefits of conventional versus smart façade systems ........................... 254

Table 7-14: Values (million$) of building component expenses/savingsby using smart façade system ............ 257

xxiv

ABSTRACT

Strong earthquakes cause severe shaking, mostly lateral, of the ground over a large area

which imparts strong excitation to building structures. These earthquakes are extreme

actions, from which buildings may not survive unless being properly designed in advance.

In recent years, many new devices such as energy-absorber or isolation systems have been

introduced. But, most of them have some disadvantages such as complexity of design and

requirement of large spaces for installation and significant cost. To date the engineering

community has seen structural facade systems as non-structural elements with a high

aesthetic value and a barrier between the outdoor and indoor environments.

As an integral part of buildings, they are susceptible to potential failure when subjected

to severe environmental forces such as earthquake and high wind in case they are not

designed properly. Seismic loads can potentially impose significant in-plane loading on

the facade system and may lead to damage and breakage in the case of insufficient

connection detailing and big inter-storey drifts. The role of facades in reducing energy use

in a building has also been recognized and the industry is witnessing the emergence of

many energy efficient facade systems. Despite these advancements, the facade has been

rarely considered or designed as a potential earthquake-induced vibration absorber for

structural buildings.

Development and implementation of advanced facade systems for enhancing the

seismic response of building structures have been a topic of debate for structural and

architectural engineers for some time. The main idea here is to design and implement a

seismic control method using a novel façade system, as an energy-absorbing device, to

decrease the level of energy imparted to the main structure during seismic activities and

xxv

hence reliance structural ductility to dissipate seismic energy. Various configurations and

specifications of the proposed system are suggested in this thesis. Multiple design

variations were evaluated as well. To prove the concept and find the optimum value of

façade damper properties, a series of non-linear structural analysis and finite element

modelling was conducted using SAP2000 and ANSYS program respectively. First,

conventional façade brackets were replaced by the so-called sacrificial elements, which

can have back and forth movements during earthquake excitations. Predefined plastic

hinge behaviour is suggested for the façade bracket elements in a double skin façade

system.

Second, façade bracket properties in terms of stiffness and damping of the proposed

system were optimized to obtain the desired response. Third, the potential of utilizing a

movable exterior skin in a double-skin facade was investigated and it was found that with

optimal allowance of façade in-plane movement and appropriate bracket stiffness, a

substantial portion of earthquake-induced vibration energy can be dissipated, which could

lead to avoiding expensive seismic designs. A series of dynamic time history analyses

were carried out to determine the behaviour and response of the proposed system on

typical concrete frame structures under different intensity earthquakes. SAP2000 and

ANSYS programs were used for the numerical analyses in all phases of the feasibility

study.

The initial works demonstrated that the seismic response for low- and mid-rise

structural buildings subjected to moderate to severe earthquakes can substantially be

reduced by the introduction of a smart design of a double skin system. Application of

flexible connections in façade systems can, if properly designed, reduce the inelastic

xxvi

deformation of structural models in comparison with the case without flexible

connections.

KEYWORDS: Façade Systems, Multi-Skin Façade, Multi-Storey Building, Seismic Load,

Wind Load

1

CHAPTER 1

INTRODUCTION

2

1.1 Background of the Study

Advances in building heights are often accompanied by increased flexibility and

inadequate inherent damping in building structures. In addition, it leads to increased

susceptibility of structures to the actions of wind and earthquake. Earthquakes have

been one of the biggest challenges for structural engineers, because they occur without

any warning and cause great damage to structural elements. Fundamental vibration

period and frequency of structures have direct correlation with their overall height.

This frequency is within the approximate range of 0.3 to 3.0 Hz for low- to mid-rise

buildings.

On the other hand, frequencies associated with most of the earthquake records are

between 1 and 5 Hz. Therefore, they affect a wide range of structures from mid- to

high-rise structures due to potential resonance compared with wind gust with lower

natural frequencies. Therefore, resonant conditions, in which the frequency of seismic

forces is similar to that of building structures, can most probably happen during

seismic activities in the case of mid-rise structures. In the case of seismic activities,

low- to mid-rise buildings may be susceptible to excessive inter-story drifts, large

shear forces, and noticeable accelerations.

Building structures designed to withstand seismic shocks have been undergoing a

critical appraisal in recent years, with the emphasis changing from "strength" to

"performance". Performance-based design and analysis are starting to dominate

research, development, and practice of earthquake engineering, particularly after the

1994 Northridge and 1995 Kobe earthquakes. Damage to structural elements, if not

designed properly based on modern codes, is very common and makes them

vulnerable to collapse during strong ground excitations. Under resonant conditions,

3

response of the structure reaches a maximum value and many structural elements may

collapse due to large overall displacements of the structure. Structural and earthquake

engineers have attempted to reduce the earthquake-induced response of main structure

using a combination of stronger materials and allowing for ductility. In addition, the

level of damping in a conventional elastic structure is relatively low; therefore, the

amount of energy dissipation during transient disturbances is also low, which is not

desirable in terms of their dynamic behaviour.

During strong excitations, such as earthquakes, conventional structures usually

deform well beyond their elastic limits and, finally, fail or collapse, because most of

the energy is absorbed by the structure itself through localized damage. Therefore,

various state-of-the-art technologies are used in the design and construction of new

buildings in order to improve their dynamic behaviour. Acceptance of innovative

systems in building structures is based on a combination of performance enhancement

versus construction costs and long-term effects. New innovative devices need to be

integrated into these structures with realistic evaluation of their performance and

influence on structural systems as well as verification of their ability for long-term

operations.

One of these approaches includes adding energy absorbers to a structure. The aim

of involving energy absorbers in a structure is to allow for hysteretic behaviours on

specially designed and detailed regions of the structure and avoid inelastic behaviour

in primary gravity load-resisting structural elements. Correctly implemented, a perfect

damper should be able to simultaneously decrease both stress and deflection in the

structure. So, increasing the overall damping ratio by devices such as damper systems

is a desirable solution in comparison with costly stiffening systems such as belt truss

4

and out riggers, which increase mass and cost of a structure to a large extent. Failure

of non-structural elements, especially façade systems, has been very common during

seismic activities. In recent decades, there has been an increasing desire for fully

glazed building structures and consequently façade panels, hence, non-structural and

other susceptible elements are in danger of damage and brakeage during earthquake

excitations.

Facade elements have been considered as non-structural elements mainly because

of their mass and stiffness in comparison to the main structure and their outer position

(Thambiratnam 2010, Tasligedik, Pampanin et al. 2012). On the other hand, protection

of structures is now shifting from complete reliance on the inelastic deformation of the

structure to dissipating the energy associated with severe dynamic loadings, through

the application of passive, semi-active, and active structural control devices to

mitigate excessive responses from dynamic excitations. Because of the significant

attention of building owners to facade elements as an aesthetically appealing feature

of structures, protection or mitigation of their damage is very crucial in terms of

economic and safety issues.

Interaction between structural and façade engineers have been rather limited in the

design of a building structure and both sides, especially structural engineers, do not

have sufficient information about structural role of façade panels (Hareer 2007),

(Moon 2009). More importantly, as architects develop new and leading edge creative

facade designs, it becomes more critical to focus on their connection details and

interaction with the primary structure. The design performance of facades as non-

structural elements has now been mainly focused on evaluating the damage sustained

by facade frames with fixed, not movable, connections. Incorporating facade systems

5

equipped with energy dissipative devices in order to damp out some of the seismic

energy during earthquake excitation and development of a new generation of façade

bracket elements are new topics which were discussed here (Behr 2006). Emphasis of

this research is on evaluating the interrelation between newly designed façade systems

and building structures in order to reduce the seismic response of main structure for

new buildings as well as retrofit of existing buildings (Goodno B. J. 1996).

1.2 Research Problem

It should be noted that loss of valuable and prime space coupled with initial cost of

installing large sized damper systems has been accepted unwillingly by building

owners and any viable alternative system to dissipate earthquake induced excitations

will be welcomed by the owners. Façade panels with various shapes, weights, and

connections usually act as a barrier between the indoor and outdoor environments.

They also play a major role in enhancing the aesthetics of building structures and are

attached by different types of connections to the primary structure or inner panels in

double skin façade systems. It has been reported that structural and non-structural

elements are likely to be impaired with different levels of damage in the case of a

severe earthquake.

Damage to façade panels is very common when subjected to environmental loads

(Hareer, Environment et al. 2006). The glazing systems, which are attached firmly to

the main structure, are not able to endure large lateral deformations, which can result

in significant dysfunction or failure at the strut connections or frame panels. Because

of the fixed attachment of façade panels, they have to move back and forth with the

primary structure during earthquakes and this movement causes damage and racking

to the façade panels and their surrounding structure (Sivanerupan, Wilson et al.

6

2008). Moreover, even small-scale damage to façade sealants has an impact on

thermal and weather insulation and leads to significant additional cost to building

maintenance. Also, with the crucial role of facades for safety of occupants and the

primary structure, only few studies have been conducted regarding the safety issues of

these systems when they are subjected to dynamic loads. Structures recently built in

seismic regions have been designed according to modern seismic codes and, in most

cases, could withstand the expected earthquakes with only a moderate amount of

damage; but, with more severe excitations, there would be much more damage

(Hareer, Environment et al. 2006).

Thus, the demand for the analysis and design of structural and non-structural

elements that can resist extreme loading has become much more important over the

past decades. The seismic analysis of glazing and facade systems, particularly their

connections, is a comparatively new field of research that needs more consideration.

Damage to façade panel(s) is a clear contributor to the overall damage and

accessibility to buildings after a seismic event. The current trend in building façade

and vulnerability of façade systems to extreme loads has led to increased demands for

safer and more economical structural façade (Saelens, Roels et al. 2003, Hareer,

Environment et al. 2006, Hareer 2007).

1.3 Research Objective and Aims

1.3.1 Research Objectives

The main objective of this research is to conduct a comprehensive analytical study

on the seismic response of movable façade systems equipped with passive damper

system and to highlight and assess the possible scenarios of new bracket connections

that can reduce damage and prevent the progressive collapse of panels' connections,

7

frame panels, and the primary structure. The major objectives of this study are as

follows:

1. In the first stage, to show the preliminary feasibility of the movable concept,

SAP2000 and ANSYS structural analysis packages were used to model two-

and three-dimensional frame structures with and without moving façade

systems. Various combinations of façade systems in terms of mass ratio,

frequency ratio, damper location, and number of dampers were analysed to

prove the feasibility of the proposed system.

2. To determine an optimal façade system, the second phase is devoted to full

scale analysis using ANSYS finite element package. The analysis determined

the sensitivity of parameters such as height of the main structure, intensity of

applied earthquakes as well as supporting stiffness of bracket elements in

achieving an optimal system. The evaluated parameters in this project are as

follows:

i. 4 Structural models (10 and 30-storey buildings in 2-D and 3-D)

ii. 4 Different earthquake records

iii. Different connection properties (selection of optimum dynamic properties

for façade brackets)

iv. Facade types and influence of mass

3. To prove the feasibility and effectiveness of the new smart façade system,

prototype testing is required as part of experimental investigations. As part of

the future works of this program, and based on the results of previous works,

Permasteelisa Group (the project industry partner) will fabricate and deliver at

least one façade unit with designated damper systems (passive or semi-active)

for testing.

8

4. To conduct a financial viability and cost benefit analysis of the proposed

system in terms of initial investment as well as life cycle costs of utilising

facade dampers instead of mechanical dampers.

1.3.2 Thesis Aims

A new facade damper system will concentrate on different types of flexible/energy

absorbing façade systems and their behaviour under earthquake loading is assessed. A

substantial number of numerical, analytical and experimental analyses related to the

applied seismic loads applied to the facade systems will be performed. The façade

system will be an energy absorbing one incorporating specially designed passive

dampers. Design of a new energy absorbing façade system, fully tested against

selected applied earthquake loads, is the ultimate goal of the research.

1.4 Methodology of Research

Main methods used to carry out investigations on the feasibility and performance of

the proposed system are based on analytical observations. The main part of this

research was conducted using computer simulations including time history nonlinear

analysis and finite element analysis for the primary and detailed modelling,

respectively. Material and dynamic properties (stiffness as well as mass and natural

frequencies of vibration) of the main structure, damping property, configuration and

location of dampers, facade mass and selected earthquakes were the influential

parameters. Based on all above parameters, properties of optimum facade connection

were evaluated. Finally, the results of investigation showed that with an innovative

and precise design of bracket façade, stack joint and structural sealants significant

reduction in structural response can be achieved.

9

1.5 Scope of Research

Effect of movable façade system as a passive damper on the dynamic behaviour of

primary structure are investigated in this research. Movable façade systems have the

potential to reduce the lateral displacements and accelerations of buildings as well as

level of damage in structural elements. In areas prone to severe earthquakes, effects

of seismic activities can be mitigated using energy absorption devices or mechanisms.

Hysteretic behaviour of bracket elements, stiffness of brackets, direction of facade

movement, and arrangement of these devices are very crucial in the reduction of

damage in structural buildings. Another major factor considered in the design

procedure is: How much additional cost is involved when dissipative smart façade

systems are used?

1.6 Dissertation Layout

This dissertation was organized in seven chapters. An introduction about the

research proposal and objectives is provided in this chapter. Literature review of the

relevant previous research in this area and general information about earthquake,

structural design, and damper systems are presented in Chapter 2. Different type of

façade systems and previous analytical and experimental studies on the interaction of

façade systems with the main structure during wind and earthquake excitations are

summarized in Chapter 3. However, the previous studies did not take a holistic

approach to movable façade systems, and have not considered façade panels as

structural elements. Thus, preliminary numerical modelling and feasibility study of the

proposed idea were developed in Chapter 4. Nonlinear analytical models of multi-

storey buildings with different cladding types and interaction between different façade

systems with main structure during earthquake excitations are presented in this

10

chapter. To demonstrate the effectiveness of the proposed system, comprehensive 3-D

dynamic time-history analyses of 10- and 30-storey structural models are presented in

Chapters 5 and 6, respectively. Financial assessment and feasibility of the proposed

system with some case studies are then discussed in detail in Chapter 7. Finally, the

main results and conclusions of the dissertation and future research needs are

summarized in Chapter 8.

11

CHAPTER 2

GENERAL INFORMATION

ABOUT EARTHQUAKE

LOADS AND METHODS OF

MITIGATING SEISMIC

ACTIVITY

12

2.1 Introduction

The race toward new building heights has not been without its challenges.

Unfortunately, these advances in height are often accompanied by increased flexibility

and shortage of adequate inherent damping in buildings. In addition, it leads to increased

susceptibility of structures to the actions of wind and earthquakes among other loads.

Average return period of earthquake excitations versus seismic hazard (measured in

terms of peak ground acceleration) is shown in Figure 2-1.

Figure 2-1: Seismic hazard versus return period (Paulay 1992)

Additionally, predicting the frequency content of earthquakes is quite complicated,

since their frequency content is a function of many factors including distance from the

epicentre and foundation soil conditions and containing several dominant frequencies

may amplify higher vibration modes in addition to the fundamental mode of vibration.

The level of damping in a conventional elastic structure is very low; therefore, the

amount of energy dissipated during transient disturbances is also very low. During

strong excitations, such as earthquakes, conventional structures are usually deformed

well beyond their elastic limits and finally fail or collapse. Therefore, most of the

imparted energy is absorbed by the structure itself through localized damage while it

13

is failing. As a response to the inherent inadequacies in the philosophy of conventional

seismic design, a number of innovative approaches have been suggested in recent

times.

2.2 Earthquake-resistant design for structural buildings

In early building design procedures, only gravity loads were considered. Generally,

there is no significant change in these loads during the life time of buildings. These

factors had considerably simplified the design process of buildings and allowed

engineers to design and construct magnificent structures before developing the scientific

basis and design methods of structures. The simplification process allowed the use of

trial and error method in designing, especially if the designer is not bound by economic

and material usage constraints.

In the new construction era, resources are limited; so, designs should be efficient. In

addition, we expect to be protected against environmental forces such as wind,

earthquakes, etc, which are not gravitational, static and single-component. For these

loads, inertia effects are important and lead to dynamic magnification and cyclic

response. Compared with gravitational loads, it is very difficult to predict the magnitude

and nature of dynamic loads. Using current approaches, wind and small earthquake

loadings are idealized in such a manner that the structure behaves in an elastic range,

while some damage is allowed at the time of medium to intensive earthquakes by

yielding of certain components.

Even a modest consideration of lateral forces in design improves the survival of

buildings although considerable improvement can be achieved by considering the actual

dynamic nature of the environmental disturbances. As a result, in dynamic terms, new

concepts of structural protection have been progressed or are at different stages of

14

progress. Many seismic design methods and construction technologies have been

improved over the years to decrease the seismic responses of buildings, bridges, and

potentially vulnerable structures (Paulay 1992) . Structural design standards impose

some regulations and criteria such as maximum top displacement and storey drifts to

evaluate the level of damage in structural elements and safety of occupants. Strong

earthquakes shake the ground severely (mostly laterally) over a large area and applies

strong excitations to building structures, with the result that buildings not designed to

withstand these forces may not survive (Paulay 1992, Elghazouli 2009).

Recently, the structures built in seismic regions are being designed according to

modern seismic codes and, in most cases, can withstand expected earthquakes with

reasonable amount of damage (Baird, Diaferia et al. 2011). However, if a more severe

excitation occurs, it can produce considerably more severe damage (De Matteis 2005).

Building survival in a large earthquake depends on the ability of its lateral resisting

system to dissipate energy hysterically while undergoing large inelastic deformations.

Concept of weak beam strong columns is considered in the design of structural

connections (Calvi et al. 2002). For earthquake-resistant structures, beams should deform

inelastically and dissipate energy with no damage to vital columns and should preferably

stay within the elastic range with minor plastic deformation.

Most of the design codes introduce four methods of analyses which are quasi-static

method, time history analysis, response spectrum analysis, and static pushover analysis

(Wilkinson and Hiley 2006). Time history analysis is the best method for design in order

to understand the response of a structural system during an earthquake. It involves

dynamic computer analysis of the structure under earthquake loading. A dynamic

analysis of a structure by the time history method includes calculating the response of a

15

structure at each increment of time when the structure base is subjected to a specific

ground-motion time history. The advantage of using this method over the linear elastic

response spectrum method is that it can be used to analyse the response of highly non-

linear structures as well. But, the downside of the analysis is that it generally requires

more computing effort and memory and usually most designers are only interested in the

maximum structural response, not necessarily the response at each time increment. The

applied ground-motion time histories should be appropriate for a specific site and have

response spectra which approximate an appropriate design. For the seismic design of

structural buildings, there are some crucial factors as discussed below:

1. Torsion: involving centre of mass and geometric centre of floors. Uneven mass

distribution which places the centre of mass outside the geometric centre and leads to

torsion (stress concentration) and should be avoided if possible or at least minimised.

It should be mentioned that a certain amount of torsion is unavoidable in every

building design. Symmetrical arrangement of masses and stiffening elements will

result in balanced stiffness and mass in orthogonal directions and keep torsion within

a manageable range.

2. Damping: Buildings in general are poor absorbers of dynamic shocks.

3. Ductility: For a given seismic demand, if the structure (or one’s design) does not

remain elastic, it undergoes plasticity/fracture/damage; then, stiffness could drop

dramatically and deformations will increase significantly. Under these increased

deformations, structural designers should ensure that the structure remains stable

without collapsing, should not loose vertical load-carrying capacity, and should be

detailing it in such a way that can undergo large deformations without collapsing.

The ability of a structure to undergo large deformations without collapsing is

called ductility and the detailing of the structure that enables it to have large ductility

16

is called ductile or ductility detailing. Normal reinforced concrete structures are

considered as non-ductile materials and fail abruptly by crumbling. Good ductility

can be achieved by carefully detailed joints. Overall ductility factor of structural

models is evaluated by comparing top lateral displacement of structure in case of

static load (elastic deformation) and earthquake (plastic deformation). For an

appropriately designed structure, this ratio should be around 2 to 4. Local ductility

for each structural element is evaluated by comparing elastic strain and plastic strain

in both columns and beams.

4. Strength and stiffness: Degree of resistance to deflection or drift (stiffness) and

ability to withstand loads safely (Strength). Property of a material to resist and bear

applied forces within a safe limit

5. Building configuration: determines the way seismic forces are distributed within the

structure, their relative magnitude, and load paths.

With well design of connections, plastic hinges should be formed in beam elements;

absorb earthquake energy, which in turn change frequency of the structure. As a result of

this change, frequency of structure would change from frequency at resonance triggered

by earthquake acceleration. Based on the aforementioned approach, order of failure in a

well-designed structure should be as follows: Beams<Connection<Columns. It is highly

recommended in seismic design to make roof floors as light as possible (light weight or

thin roofs) and ensure towel designed connections to hold the whole frame together

(strong moment connections). There are two major philosophies in the design of

structural buildings, as discussed below.

17

2.2.1 Force Method

Structural design philosophy has been, until recently, focused on forced-based (FB)

design methods, in which inertia of the structure generates forces within the structure.

This method uses over-strength and ductility factor (or structural response factor) to

evaluate the performance of structure (Chandler and Mendis 2000).

2.2.2 Displacement Method

In this method, a limit is placed on inter-storey drifts and overall building lateral

deflection according to displacement-based design. In this method, design of building

structures should be according to top displacement limitation and inter-storey drift which

should be less than and 1.5% of storey height, respectively (Chandler and Mendis

2000). This method requires the structure to be represented as a single-degree-of-

freedom structure and the seismic performance is assessed by comparing the

displacement demand with the estimated structural displacement capacity (Fajfar 2000).

The displacement capacity (∆ ) is obtained from a non-linear push-over analysis where

the designer calculates displacement as a function of increasing horizontal force until the

structure is deemed to have failed (Ghobarah 2001). Structural failure is assumed to have

occurred when the overall structure ceases to be able to support gravitational loads and

collapse is followed (Priestley 2000). The resultant force-displacement plot is commonly

known as the "push-over" (or capacity) curve which indicates the capacity of the

structure to deform (Doherty, Griffith et al. 2002). Calculations for developing the

transformed capacity curve are material-dependent, but should include effects such as

elastic and inelastic deflections of the structure together with deflection contributions

from foundation flexibility and P-delta effects (Smith, Coull et al. 1991).

18

2.2.3 Review of AS1170.4 Australian Standard

Australian Standard 1170.4 requires earthquake analysis for all buildings and utilizes

a three-tiered approach, depending on earthquake design category (EDC):

EDC1 – Simple static analysis (10% weight of each floor is applied to each floor of

the structure)

EDC2 – Static earthquake analysis

EDC3 – Dynamic earthquake analysis

Most designers use the force-based principles of EDC1 or EDC2, except in the design

of tall buildings (where higher mode effects are important) in which EDC3 method is

often used. The new standard also allows the designer to undertake a displacement-based

check for earthquake code compliance following a design for gravity and wind loads,

which is often sufficient in low-seismicity areas on rock or firm soil sites. There is an

important distinction between this definition of failure (in terms of ensuring sustained

gravitational load-carrying capacity) with the traditional definition of failure used in high

seismic regions for ensuring that horizontal resistance capacity is at least 80% of the

nominal capacity (AS1170.1 2002, NZS1170.5 2004)

2.2.4 Seismic retrofitting of building structure

Seismic retrofitting refers to modification of the existing structures to make them

more resistant to seismic activity, ground motion, or soil failure due to earthquakes and

hence code compliant. Different retrofitting strategies and technical methods have been

used by structural designers to achieve an overall retrofit performance objective, such as

increasing strength, increasing deformability, and reducing deformation demands (Ma,

Cooper et al. 2012). Design of structures with adequate detailing and reinforcement for

19

seismic protection has additional cost for building owners as well as requiring increased

time of construction. Moreover, retrofitting the existing structures which are old and/or

non-engineered is too expensive and building owners may prefer to demolish the whole

structure and start again in many developing countries (Martinez-Rodrigo and Romero

2003). It is also crucial to keep in mind that there is no such thing as a fully earthquake-

proof structure; however, seismic performance can be significantly enhanced through

proper initial design, subsequent rehabilitation, or modifications using novel as well as

traditional ideas (Asadi, Da Silva et al. 2012). A new way of retrofitting an existing

building structures uses façade panels which have been considered as non-structural

elements thus far and this is the main thrust of this thesis.

2.3 Drawbacks of Current Non-structural Australian and international

Standards for the Design and Damage Assessment of Primary Structure

and Façade Panels

2.3.1 Introduction

The role, played by the so-called "non-structural components", is not incorporated

into either Australian or international standards and is therefore, not considered in the

current design process (Behr et al. 2006). Current Australian Standards 4284, 1288,

2047 and FEMA356, 389 for the damage assessment of non-structural components are

discussed and any necessary revisions needed for new bracket connections are

proposed in this section. It is costly to design and construct a building to withstand the

maximum probable earthquake; therefore, the designer and owner must establish an

acceptable level of damage at an appropriate risk level. This acceptable level of risk

should include the degree in which the glazing unit will can be damaged and the level

at which there will exist the possibility of air and water penetration into the building,

because structural and non-structural performance levels are not necessarily the same

20

(AS1288.4 2006, Behr 2006). Non-structural parts, including their components and

fasteners, should be designed for horizontal and vertical earthquake forces, because it

is mainly anticipated that the level of damage to non-structural components would be

worse than that of the main structure (FEMA356 2000, Behr 2006) . The current

seismic design provisions typically require non-structural components to be secured so

that they do not present a falling hazard; but, they may still undergo so severe damage

that they cannot function any longer (Palermo, Pampanin et al. 2010) .The major

concern regarding glazed facade panels is racking action due to the relative lateral

movement of buildings from earthquake excitation. Some methods for improving

these components, such as adding smooth corners around each glass panel and using

heat strengthened, toughened, and laminated glasses, have already been suggested;

but, at present, the seismic drift performance of glazed facades is not considered at the

design stage by facade engineers (Behr 1998).

Australian Standard AS4284 is more concerned with testing building facades for

determining the performance of a representative building facade under simulated

loading conditions before production commences, rather than using a sample as an

opportunity for the manufacturer to evaluate the fabrication and installation process

(AS/NZS4284 2008). This standard is applied to all types of facades, including low- to

high-rise, commercial, industrial, and residential buildings with fixed facade

connections. According to the standard's recommendation, a seismic loading test is

necessary and applicable for testing prototypes in a test facility and on-site testing.

The parts and components of facades should be designed for earthquake actions using

the established principles of structural dynamics, general method using design

accelerations, or the simplified method. The first method implies an accurate dynamic

analysis, which is time-consuming and tedious; but, the second and third methods are

21

more practical and will be described and discussed below. Based on Section 8 of AS

1170.4, the architectural components and their fixings should be designed for the

acceleration determined by the design methods given in this standard, while the force

generated on the facade components can be calculated from equation (2-1) (AS1170.4

2007)

0.5 (2-1)

Based on a simple method, panels should be designed to resist the force of a

horizontal earthquake as determined in Equation 2.2. The force should be applied to

the centre of the mass in combination with the weight of the element.

0 (2-2)

= Effective acceleration of the floor at the level where the panel is situated,

calculated from the earthquake actions determined for the structure using other

sections of the standard

Component importance factor taken as:

1.5 for the components critical for life safety, parts required to function

immediately following an earthquake, and all the components in a structure which are

at importance level 4

1 for all other components

Component amplification factor

2.5 for flexible spring-type mounting systems for mechanical equipment

=1 for all other mounting systems

Component ductility factor

22

=1 for rigid components with non-ductile or brittle materials or connections

= 2.5 for all other components

Seismic weight of component (Knaack, Klein et al.)

Probability factor

Hazard factor

Height amplification factor at height where the component is attached

0 Bracketed value of the spectral shape factor for the period of zero sec

It should be noted that values for all mentioned factors can be calculated from

AS.1170.4. Based on FEMA 356, the facade frame is designed in accordance with the

capacity design principles, while the system is designed according to the drift limits in

Serviceability Limit State (SLS) and the Ultimate Limit State (Traulsen and

McClellan). The serviceability limit state (SLS) is specified in the form of deflection

limits for earthquakes by design standards.

These deflection limits are related to earthquake actions with annual probability of

exceedance of 0.04 (Behr, Belarbi et al. 1995, FEMA356 2000) ; i.e. corresponding to

a return period of 25 years. The ultimate limit state (ULS requirement is defined such

that the facade should remain supported and not interfere with evacuation for the

design level earthquake; also, facade damage is expected in an ULS event according

to the current design standards, because the SLS limits define deflections beyond

which repairs can be expected. However, the damage should not be life-threatening.

For a safe design, the panels' drift should be between the range of SLS and ULS; but,

even with larger deflections than the SLS limit, a smart and innovative design for the

facade brackets can reduce the risk of extreme damage to the facade panels (Behr,

23

Belarbi et al. 1995, FEMA356 2000). Glazing should satisfy the design requirements

for ultimate and serviceability limit states in accordance with the procedure given in

the standard. The standard for earthquake actions in Australia is AS 1170.4 -2007; it

limits the inter-story drift to 1.5% of storey height in buildings and also proposes that

attachment of cladding and facade panels to aseismic-force-resisting system should

have sufficient deformation and rotational capacity (AS1170.4 2007,

Sivagnanasundram 2011) .

However, with rigid connections and large earthquake forces, it is very difficult and

risky to allow large deformations and apply inertia forces without incurring much

damage in the brackets, facade frame, and main facade frame. This means that

moveable brackets with flexible materials or newly designed attachments are the best

ways for reducing the risk of damage and threats to occupants and passing pedestrians

(Rajgopal and Jayachandran 2012) . Performance-based seismic design is a relatively

new concept that reflects a natural evolution in engineering design practice. It is based

on the investigations of building performance in past earthquakes and laboratory

research and has resulted in improved analytical tools and computational capabilities.

Building performance can be described qualitatively in terms of (FEMA356 2000):

Safety afforded to building occupants during and after an earthquake;

Cost and feasibility of restoring the building to pre-earthquake conditions;

Length of time the building is removed from service to conduct repairs;

Economic, architectural, or historic impacts on the community at large.

The primary function of performance-based seismic design is the ability to achieve,

through analytical tools, a building design that will reliably perform in a prescribed

manner under one or more seismic hazard conditions. The performance, or condition

of the building as a whole, is typically expressed through qualitative terms and

24

intended to be meaningful to the general public. These terms use general terminology

and concepts describing the status of the facility (i.e. Fully Operational, Operational,

Life Safety and Near Collapse), but should be also associated and linked to

appropriate technically-sound engineering terms and parameters. These performance-

based design criteria are applicable to both structural and non-structural elements.

Older infill panels do not contain any in-plane movement allowance apart from the

small gaps around each panel of glass. These gaps are typically only a few millimetres

wide and consequently only allow a minimal amount of in-plane drift before the glass

begins to carry force.

The building performance levels typically are the structural performance level that

describes the limiting damage state of the structural systems, plus a non-structural

performance level that describes the limiting damage state of the non-structural

systems and components. These performance characteristics are directly related to the

extent of damage sustained by the building during a damaging earthquake. The

maximum in-plane drift demand and performance assessment of facade panels have

already been described in some books on design, and also by previous researchers.

The level of damage to each facade is called facade performance levels (or damage

states), suggested by FEMA356, and is determined in terms of the following

performance levels shown in Figure 2-2(FEMA356 2000).

1. Operational performance level 2. Immediate Occupancy performance level 3. Life Safety performance level 4. High Hazard performance level or Collapse Prevention

25

(a) Operational (b) Immediate Occupancy (c) Life Safety (d) Collapse Prevention

Figure 2-2: Seismic performance levels of a building

Table 2-1 describes the damage associated with the four non-structural

Performance Levels of Collapse Prevention, Life Safety, Immediate Occupancy, and

Operational for specific types of architectural components (FEMA356 2000). TABLE

II from FEMA 389 Chapter 4 illustrates this concept comprehensively.

Table 2-1: Non-structural performance level

Component

Non-structural Performance Level Collapse

Prevention Level

Life Safety Immediate Occupancy Operational

Cladding

Severe damage to

connections and cladding. Many panels

loosened

Severe distortion in connections. Distributed

cracking bending, crushing and spalling of cladding elements. Some fracturing of cladding but

panels do not fall

Connections yield minor

cracks(<1/16" width) or

bending in cladding

Connections yield minor

cracks (<1/16"width) or

bending in cladding

Glazing

General shattered glass and distorted

frames, widespread

falling hazards

Extensive cracked glass, little broken glass

Some cracked

panes; none broken

Some cracked panes; none

broken

Light Fixtures

Extensive damage,

falling hazards occur

Many broken light fixtures. Falling hazards

generally avoided in heavier fixtures

Minor damage.

Some pendant lights

broken

Negligible damage

A Target Building Performance Level is designated by the number corresponding to

the Structural Performance Level (identified as S-1 through S-6) and the letter

26

corresponding to the Non-structural Performance Level (identified as N-A through N-

E). Note that the four Target Building Performance Levels discussed above are each

designated as follows:

Operational Level (1-A): Immediate Occupancy Structural Performance Level (S-

1) plus Operational Non-structural Performance Level (N-A).

Immediate Occupancy Level (1-B): Immediate Occupancy

Structural Performance Level (S-1) plus Immediate Occupancy Non-structural

Performance Level (N-B).

Life Safety Level (3-C): Life Safety Structural Performance Level (S-3) plus Life

Safety Non-structural Performance Level (N-C).

Collapse Prevention Level (5-E): Collapse Prevention Structural Performance

Level (S-5) plus Not Considered Non-structural Performance Level (N-E).

The seismic rehabilitation of the existing architectural components that is

permanently installed or is an integral part of a building system has been suggested by

FEMA. An assessment process is needed to determine which non-structural

components are to be rehabilitated. These rehabilitation requirements are related to the

zone of seismicity and Hazards Reduced, Life Safety, and Immediate Occupancy

Performance Levels. Based on the FEMA 356 recommendation, non-structural

components should be rehabilitated by completing the following steps:

1. The rehabilitation objectives shall be established by selecting a non-structural

Performance Level and earthquake hazard level in accordance with FEMA

Section11.4. The zone of seismicity shall be determined as well.

2. The components shall be assessed by inspecting them.

3. Analysis and rehabilitation requirements of the component for the selected non-

structural Performance Level and appropriate zone of seismicity shall be determined.

27

4. Interaction between structural and non-structural components.

5. Each type of non-structural component shall be classified.

6. Acceptability of connections between non-structural components and the

structure shall be established.

7. If non-structural components which do not meet the requirements of the selected

non-structural Performance Level they shall be rehabilitated (FEMA-389 2004).

FEMA recommends that a direct visual inspection of each type of non-structural

component in the building shall be carried out. They shall be assessed by checking the

presence and configuration of each type of non-structural component, their attachment

to the structure, their physical condition, and whether or not degradation is present and

whether the non-structural components could potentially influence the overall

performance of the building (FEMA356 2000). If detailed drawings of the facade

system are available, at least one sample of each type of non-structural components

shall be observed , but if there are no deviations from the drawings the sample shall be

considered representative of the installed conditions (FEMA-389 2004).

If there are deviations, then, at least 10% of all occurrences of the component shall

be observed. If detailed drawings are not available, at least three samples of each type

of non-structural component shall be observed, but if there are no deviations between

the three components, the sample shall be considered representative of installed

conditions. If there are deviations, then at least 20% of all occurrences of the

component shall be observed. Current facade standards are focused specifically on

evaluating facade systems with fixed connectors, but, specific test requirements are

necessary if a movable bracket concept is to be introduced into the market. This gap in

the facade standards must be filled as soon as the concept has been proven by

experimental tests. The impact of performance-based strategies on future design codes

28

for novel non-structural elements and their attachments should also be considered.

While current design codes clearly require life safety design for only a single level of

ground motion, the next generation of performance-based seismic design guidelines

shall provide engineers with more comprehensive guidelines to design and construct

buildings that pass a number of performance criteria when subjected to earthquakes

with varying degrees of severity. Performance-based strategies are incorporated by

using three “Seismic Use Groups.” These groups are categorised based on the

occupancy of the structures and the relative consequences of earthquake-induced

damage to the structures as follows:

Group I structure such as low-rise commercial office buildings with basically lower life hazards

Group II structures such as elementary schools with a large number of occupants Group III structures such as medical facilities that are essential for post-earthquake

recovery

The design criteria for each group are intended to produce specific types of

performance in design earthquake events, based on the importance of reducing

structural damage and improving life safety. It is expected that the successful

development of a guideline for both structural and non-structural elements will require

an enormous effort in terms of financial and technical participation (FEMA-389 2004)

Seismic design needs to be considered in order to select appropriate glazing materials which are listed as below:

1. Flexible frames to accommodate rocking without damage or serviceability issues

2. Adequate glass to frame clearances

3. Laminated glass in annealed or heat-strengthened constructions, either monolithic or in an insulating glass unit

4. Bottom and side blocks

5. Silicone glazing

29

The following items should be considered when glazing material and frame

types are going to be selected:

Energy conservation

Sound reduction qualities

Potential threats and vulnerabilities as determined by building owners, security consultants, and project design teams

Primary consideration of environmental threats will minimize budget concerns and

decrease project completion time while modifying safety protection to the highest

level. Protecting building inhabitants is another factor that should be considered. The

number of damaged frames can be reduced through the use of properly designed

movable façade systems (W. Bush, Steinberg et al. 2004).

2.4 Modifications in Structural Systems

An efficient structural system can provide the most effective means of controlling

structural dynamic response. The use of space frame and mega-frame concepts,

outrigger trusses, belt trusses and Band-Aid type stiffening systems can offer

additional resistance to earthquake and wind loads (Kareem 1992). Other alternatives

include modification of the structural mode shapes to increase the mass participating

in the dynamics of building in the fundamental mode. However, in this situation care

must be exercised as the contributing loading in the fundamental loading may also

experience an increase. Other options may include shifting the major stiffness axes

from the principal geometric axes (Kareem 1992). These solutions are however costly

with little possibility of retrofitting a structure through such modifications. Although

other conventional technics like mass damper and tunned damper are likely to be more

feasible.

30

2.4.1 Cladding Isolation

Kareem (1992)) proposed the concept of isolation in the mountings of the cladding

to the structural system. Buildings are isolated from earthquake excitation by

employing isolator bearings between the building and the foundations and a similar

concept is proposed for cladding. The integrated effects of the unsteady aerodynamic

loads acting on cladding are transferred to the frame which results in building motion.

If the cladding is connected to the frame by an isolation mounting, then the

aerodynamic loads transferred to the frame will be reduced and consequently the

building motion will also be reduced. In order for this mounting to be effective, the

ratio of excitation frequency to the natural frequency of the cladding system should be

greater than the square root of two (Kareem 1992).

In this situation the mounting system is more effective without any damping based

on principles of vibration isolation. The proposed system can be materialized by

dividing the claddings on the building envelop into several segments. The preliminary

calculations suggested that such a mounting system would be quite soft and pneumatic

mounts may be an appropriate choice. Such an installation may cause the cost of a

cladding system to rise significantly. This can be overcome by using these systems in

staggered configurations and the remaining portions of the building envelop may

utilize conventional cladding. The staggered arrangement has been proposed to help

reduce the correlation of wind-induced pressure which in turn would result in

lessening of the integrated loads (Kareem 1992)..

2.4.2 Addition of Damping Systems

Damping is becoming a part of the structural engineer’s design vocabulary. It is

well known that the introduction of damping into structural systems is a very efficient

31

method in reducing the effects of dynamic loads on these systems. Over the past

thirty-five years the idea of introducing a separate system to increase the damping in

buildings has gained widespread acceptance (McNamara, Boggs et al. 1997). Wind

engineers have introduced damping systems in large scale structures such as the

World Trade Centre (Mahmoodi, Robertson et al. 1987) and Citicorp Center

(McNamara and J 1977) in New York city. In the design of tall buildings, engineers

must assume a level of natural damping in the structure in order to assess the building

habitability during frequent wind storms. The actual damping in building structures is

a difficult quantity to measure prior to building construction and varies with response

levels, type of structural systems, cladding system and materials used for construction.

Recognizing this uncertainty associated with estimating the natural damping in

structural systems, engineers have introduced energy dissipating systems into the

design of buildings to augment damping.

These systems are designed to provide specific amounts of damping and by

controlling the damping provided the uncertainty associated with assuming the

damping present is eliminated. Structural designers attempting to solve motion

comfort problems in tall buildings have found that direct addition of damping to the

structure is the most reliable way of assuring a well-behaved structure in turbulent

environments (McNamara, Boggs et al. 1997). The use of energy dissipating systems

related to wind effects on buildings is focused on the reduction of the acceleration

response of the upper floors of a tall building. Occupant discomfort due to wind-

induced motion is strongly dependent on the turbulent and buffeting characteristics of

the wind and presently there exists no satisfactory computational procedure to

determine these effects (McNamara, Boggs et al. 1997). The dynamic characteristics

of the structure are calculated (eg vibration period, mass) and an estimate of the

32

natural damping is made based on the type of the lateral load resisting system and the

materials used in construction. Predictions of acceleration response levels for various

assumed damping values can be generated by wind engineers. A level of damping is

then selected which satisfies the appropriate design criteria. The parametric

adjustments in the structural properties; mass, stiffness, and damping can change the

overall structural response. The acceleration response of a building is given by the

dynamic equation of motion

Acceleration (2-3)

From Equation (2-3), when the stiffness is increased to reduce acceleration, the

ratio of stiffness to mass ratio, governing the natural frequency of building, should be

considered in the structural design. Slope of the load spectrum with frequency should

also be considered. Addition of structural damping or energy dissipation devices, such

as passive energy absorbers and active feedback control systems to the basic structural

system of buildings, results in the reduction of acceleration.

2.5 Energy Dissipation Systems

By adding damper devices to the main structure, the majority of energy input that is

applied to the buildings during seismic excitation can be dissipated. Therefore, structural

response can be reduced by 40-60% in comparison to the traditional structures without an

energy dissipation system and therefore the against failure increases (Aiken 1999).

Mechanical dampers can dissipate majority major proportion of vibration energy input

applied to the buildings during earthquakes or winds. They can also be installed on the

bracings, walls, joints, connection parts, non-structural elements, or any other appropriate

space in structures. Energy balance equation during an earthquake in building structures

equipped with the energy dissipation system is as follows (Aiken 1999):

33

EIin = Ep+ Ek+ Ep+ Eb (2-4)

where:

EIin: Energy input to the structure,

Ep: Potential energy during structural vibration,

Ek: Kinetic energy during structural vibration,

Ed: Energy dissipated by viscous damping of structure or equipment,

Eb: Energy dissipated by energy dissipation system,

The goal is to increase Eb so that, for a given EI, the elastic strain energy in the

structure is minimized. It means that the structure would endure smaller deformations for

a given level of input energy than if it did not contain energy dissipation system (Aiken

1999). Alternatively, increasing Eb permits Ep to be reduced for a higher level of EI. The

role of a passive energy dissipation system is to increase the hysteretic damping in the

structure. Tests and research have shown that energy dissipation systems can dissipate

about 90% of the total energy input at the end of an earthquake (Uang and Bertero 1990)

2.5.1 Passive Controllers

A passive control system does not require any external power for operation and

performance; it utilizes the motion of the structure to generate control forces (Spencer

Jr and Nagarajaiah 2003). Systems in this category are generally reliable and have low

maintenance requirements, since they are not influenced by power outages which are

common during earthquakes. Since they do not infuse energy into the system, they are

capable of stabilizing the structural motion. Control forces are generated as a function

of the response of the structure at the location of the passive control system. These

systems increase the energy dissipation capacity of a structure through localized,

discrete energy dissipation devices (Thambiratnam 2010). New civil engineering

34

structures tend to be lighter as well as more slender and have smaller natural damping

capacity than those of their older counterparts. This trend has increased the importance

of damping technology to lessen the impact or intensity of earthquake and wind-

induced vibrations (Shiba, Mase et al. 1998).

Figure 2-3: Relationship between excitation and role of passive controllers in structure

Loop (Symans 1999)

The passive systems increase the energy dissipation capacity of a structure through

localized, discrete energy dissipation devices located either within a seismic isolation

system or over the height of the structure (Kitagawa and Midorikawa 1998).

Depending on their construction, these systems may also increase the stiffness and

strength of the structure to which they are connected. Role of passive controllers in a

structure is shown in Figure 2-3. Generally, they are categorized as the following

systems (Zhou and Xian 2001).

2.5.1.1 Tuned Mass Damper

Tuned mass dampers (TMDs) have been found effective in decreasing the response of

structures subjected to dynamic loads. These devices are sub-classified based on their

mechanism of energy dissipation and system requirement (Aldemir 2003). It is important

to note that a passive TMD can only be tuned to a single structural frequency. While the

first-mode response of an MDOF structure with the implementation of TMDs can be

substantially reduced, the higher mode response may in fact increase as the number of

stories increases (Sadek, Mohraz et al. 1997). Passive TMDs are a very efficient solution

35

for the control of vibrations in the structures subjected to long-duration, narrow-band

excitations (Soong and Spencer 2002). These dampers are usually mounted on the top

floor of buildings and improve structure responses by changing their dynamic behaviour.

Using TMDs, small reduction in cumulative yielding of components is observed, whereas

the reduction in ductility ratio is non-significant (Soong and Dargush 1997, Soong and

Spencer 2002).The natural frequency of the TMD is tuned in resonance to one of the

dominant frequencies of the structure (usually the first natural frequency). The mass of

the TMD is about 2.5 - 4% of the building mass in the fundamental mode and are

mounted in the location in which the structure experiences maximum response, mostly

near the top. Space restrictions will not permit traditional TMD arrangement, leading to

adopting alternative arrangements, including multi-stage pendulums, to be used (Ji, Moon

et al. 2005).

Despite the fact that TMDs are often effective, better performance has been noted

through the use of multiple TMDs composed of several dampers placed in parallel

configuration with natural frequencies around the optimum frequency (Li 2002). TMDs

have advantages including simplicity, reliability, effectiveness, and low-cost in

manufacturing and maintenance. Their main disadvantage is the sensitivity to tuning ratio.

The mistuning or non-optimum damping will significantly reduce the effectiveness of the

TMD (Lukkunaprasit and Wanitkorkul 2001). Effectiveness of a TMD on the response of

a single-degree-of-freedom system could simply be extended to continuous structures

such as tall buildings by a modal approach. A single TMD cannot provide vibration

control for more than one mode; therefore, multi-tuned mass dampers (MTMDs) are used

to evaluate and control the vibration of multiple structural modes (Li 2002) .The

parameter optimization of a TMD is a significant issue. The parameters which directly

affect the response of the main system are mass, damping, and stiffness of a TMD.

36

It is conventional to optimize the damping and tuning frequency ratio for the purpose

of computational convenience in most studies (Sadek, Mohraz et al. 1997). They are

frequently implemented with a frictional or hydraulic component that turns mechanical

kinetic energy into heat, like an automotive shock absorber (Li 2000). Typically, dampers

are huge concrete or steel blocks mounted inside skyscrapers or other structures and

moved in the opposite direction to the structure by means of springs, fluids, or pendulum

action. A typical TMD is depicted in Figure 2-4.

Figure 2-4: Schematic View of Displacement of TMD (Soong and Spencer 2002)

2.5.1.1.1 Equation of Motion

In the first step, the response of a single-degree-of-freedom (SDOF) structure-TMD

system subjected to a vibratory force, P is considered, as shown in Figure 2-5 and then the

TMD - multi-degree-of-freedom (MDOF) system is pursued, as demonstrated in Figure

2-6 (Tsai and Lin 1993). According to Figure 2-5, the system is a SDOF system with a

TMD that acts together as a two-degree-of-freedom system. Index d represents the TMD

parameters. The governing equations are as follows:

mmm (2-5)

The equation of motion of the main system is shown in equation (2-6):

1 m u 2ξ ωu ω u mu (2-6)

Direction of Motion

Support

Floor

37

ω

c 2ξ ωm

The equation of motion of TMD is shown in equation (2-4):

u 2ξ ω u ω u u

ωkm

Cd =2ξdωdmd

(2-7)

Where:

: The damper mass to the main mass ratio

ω: The natural frequency of the main mass

C : Main system damping:

: The natural frequency of the damper mass

: Absorber damping

ξd : Damping ratio of TMD

ξs : Damping ratio of the main mass

Figure 2-5: Single Degree of Freedom System with the Use of TMD (Soong and Dargush

1997)

The main aim of adding a TMD is to limit the movement of the main structure when it

is affected by a particular excitation. The design of a TMD is to specify mass md, stiffness

38

coefficient Kd, and damping coefficient Cd. The governing equation can be simplified in

the following matrix form:

(2-8)

Stiffness and damping matrices are diagonal; but, mass matrix is non-diagonal. This is

because of the relativity of second-degree-of-freedom (ud) to the first-degree-of-freedom

(u). This simple model is very useful in the field of vibration control of structures. The

MDOF system of Figure 2-6 is a building with a TMD that is mounted on the last floor

and the equation of motion of a structure with n degrees-of-freedom (DOF) subjected to

external disturbance, w (t), can be described as:

Msxs(t)+Csxs(t)+ksxs(t)=EsW(t) (2-9)

Where:

Ms, Cs, ks are n*n mass, damping and stiffness matrices of the main structure, respectively

xs(t) : n*1 displacement vector

W(t) : q*1 external disturbance vector,and q is the number of disturbances

Es : n*q location matrix of external disturbances

In reality, a TMD system is aliged according to the main Axes of the local floor (for

example, if the TMD is aligned in longitudinal direction x, it moves only in this

orientation). In normal structures, the impacts of torsion on the movement of a TMD can

be ignored because of its very small value. Equation (2-9) can be expanded in matrix form

for a system with one degree-of-freedom (at each floor), as follows:

M oo M

uu

C OO O

uu

K Oo O

uu + (2-10)

39

⋮, ,

, ,

⋮, ,

, ,

Where:

M ∗ MassMatrixwithoutTMD

Mm , oo o MatrixofTMDmass

C = Damping matrix without TMD K = Stiffness matrix without TMD

u ∗

u ,

⋮u ,

⋮u ,

, u u , , u u , , r ∗

100⋮100

Figure 2-6: Multi-Degree-of-Freedom System with TMD

X

Y

40

2.5.1.1.2 Determining TMD Parameters

In the first step, in designing the TMD mass, frequency and damping coefficient must

be defined. Mass of a TMD is a percent of the first modal mass (or the mode that the

TMD is tuned to control it) and the frequency of motion of a TMD is percentage of the

main frequency of the same mode (Clark 1988). Choice of damping coefficient is

somewhat arbitrary; but, it must be realistic and practical. If T is the first mode period of a

building, a TMD, with a natural period in the range of T- ε to T+ε, is chosen in order to

remain effective. The optimal damping ratio of TMD is adopted from Den Hartog's

recommendations (Rana and Soong 1998):

∗ ∑

∑ ,

MTMD μ ∗

ωTMD=γfω

KTMD=MTMDω TMD

CTMD=2ζTMDMTMDωTMD

(2-11)

CoefficientsM,γ , and ζTMD must be defined beforehand. The last step in designing the

TMD is to check and verify that TMD provides responses which are within the acceptable

responses of the building; otherwise, the design process is repeated.

2.5.1.2 Tuned Liquid dampers

Another idea to reduce the vibration of the primary structure is to use the tuned

liquid dampers (TLDs), as shown in Figure 2-7, Tuned liquid dampers, encompassing

both tuned sloshing dampers (TSDs) and tuned liquid column dampers (TLCDs), have

41

become a popular inertial damping device (Fujino et al., 1992, Kareem and Sun, 1987,

Kareem, 1990, Modi and Welt, 1987, Sakai et al., 1989, Tamura et al., 1988) since

their first applications to ground structures in the 1980s. In particular, TSDs, which

are extremely practical, are currently being proposed to double up with existing water

tanks on buildings (Fujino, Sun et al. 1992). By simply configuring internal partitions

into multiple dampers, the tanks may be utilized as auxiliary damping devices without

adversely affecting the functional use of the water supply tanks(Mondal, Nimmala et

al. 2014). The shallow water configuration dissipates energy through viscous actions

and wave breaking.

Figure 2-7 Illustration of a schematic model of a TLD

While the natural frequency of a TLD may be simply adjusted by the depth of

water and dimension of the container, there are practical limitations on the water depth

and thus the frequency which may be obtained by a given container design. Usages of

TLDs in full scale are shown in Figure 2-8. TLDs have also been used for suppressing

wind-induced vibrations of tall structures (Reed, Yu et al. 1998). In comparison with

TMDs, the advantages associated with TLDs include low initial cost, virtually free of

maintenance, and ease of frequency tuning (Symans 1999, Hemalatha and Jaya 2008).

42

Figure 2-8: TLD used in Rincon Hill (the first U.S. residential tower) (Soong and Spencer 2002)

Each TLCD has a broad horizontal chamber at the bottom with a column of water

at each end. The dampers work by allowing the water to move back and forth along

the bottom chamber of the tank and up into the columns (Soong and Dargush 1997,

Soong and Spencer 2002).

2.5.1.3 Multiple Tuned Mass Dampers

In order to overcome the aforementioned limitations in a TMD, multiple Tuned

Mass Dampers (MTMDs) were proposed, analyzed and tested. As its name suggests,

MTMDs consists of a number of TMDs which are attached to the primary structure in

parallel as illustrated in Figure 2-9, or in series as illustrated in Figure 2-10. Hereby,

the MTMDs in parallel are introduced first.

Figure 2-9 Schematic model of multiple TMD (MTMDs) in parallel

43

Figure 2-10 Schematic model of multiple TMD (MTMDs) in series

Tuned Mass Dampers (2TMD) to attenuate the structural responses under harmonic

excitations were proposed (Iwanami and Seto 1984). Optimum design for the 2TMD

was established by the authors and its better robustness than a single TMD was

demonstrated. However, the improvement of the robustness does not seem to be

significant enough. Other researchers first proposed the application of Multiple Tuned

Mass Dampers (MTMDs) with distributed natural frequencies over a given frequency

range (Igusa et al. 1990, 1991). The dynamic system is subjected to random

excitations. Yamaguchi et al. (Yamaguchi et al. 1993) established the explicit formula

for the impedance of the MTMD based on an asymptotic analysis method. It was

found that the optimal design of the MTMDs has distributed frequencies centred

around the natural frequency of the primary structure. The MTMDs were

demonstrated to be more effective and robust than a single TMD with equal total

mass. Further analysis of the effect of the substructure on the response of the primary

structure where the substructures have equal stiffness and equally spaced natural

frequencies were carried out (Igusa et al. 1992). It was summarized that the behaviour

of the multiple sub-oscillator has limited performance when the number of the sub-

44

oscillators becomes large and the natural frequency becomes closely spaced. The

behaviour of the multiple sub-oscillator can be represented by an equivalent damping

when the natural frequencies span a sufficiently wide range. It was demonstrated that

the effectiveness of the sub-oscillators is more significant than that of an equivalent

single TMD when the damping of the primary structure is limited to low values. The

fundamental characteristics of the MTMDs under harmonic loadings was investigated

(Yamaguchi 1993). Analytical steady-state solutions for the primary structure and

each of the MTMD were obtained. Effectiveness and robustness of the design

parameters (frequency range, damping ratio, number of TMDs) were evaluated

numerically based on the analytical solution. The results confirmed the advantage of a

MTMD over a single TMD in reduction effectiveness and robustness.

Furthermore, the authors revealed that there exists an optimum MTMD for the

given total number of TMDs with the optimum frequency range and the optimum

damping ratio. Abe and Fujino (Abe et al. 1994) studied the modal properties of the

MTMD-structure system and the effectiveness of the MTMD. Closed-form solutions

for the modal properties (modal frequencies, damping and shapes) of the MTMD-

structure system were derived using perturbation technique. Based on the closed-form

solutions, the reduction effectiveness was examined through evaluating the equivalent

damping added to the primary structure from the MTMD, revealing that an optimum

damping ratio exists for the MTMD. The authors also proposed a critical bandwidth of

the natural frequencies for the MTMD to make multiple tuning for the MTMD. Based

on the overall results, a general design procedure regarding the mass ratio, the number

of TMD, the damping ratio is summarized in the paper.

45

Abe and Igusa (Abe et al. 1995) further analyzed the performance of the MTMD

used in structures with multiple vibration modes. Analytical results were obtained

based on perturbation theory. It was illustrated that for structures with widely spaced

natural frequencies, the response can be approximated by the response of the well-

known single-mode structure/TMD system. In comparison, for structures with “p”

closely spaced natural frequencies, at least “p” TMDs are needed to control the “p”

closely spaced modes. In addition, the placement of the TMDs in the case of structures

with “p” closely-spaced modes is demonstrated to be important. When the TMDs are

placed inappropriately, their effectiveness will be limited. It was also found in the

paper that the coupling of the closely spaced modes can be reduced by using certain

TMD parameters and placements. The original system can be approximately

represented by a set of decoupled SDOF structure/TMD systems.

The dynamic characteristics and effectiveness of the MTMDs under random

excitations which were represented by wind and seismic loadings were evaluated

(Kareem et al. 1995). Qualitatively similar findings were obtained as that concluded

under harmonic loadings. Furthermore, it was found that an optimum MTMD exists

when the frequency range, total number of TMDs and damping ratio are selected

optimally. It was also revealed that the MTMD with variable mass dampers or

variable frequency spacing alone, or the combination thereof did not show any distinct

advantage or disadvantage over uniformly distributed mass or frequency MTMD

system. In addition, the frequency range was found to be the most important parameter

in designing a MTMD, then comes the damping ratio and the number of MTMD.

Several research on the effectiveness and robustness of a MTMD under harmonic

ground acceleration were conducted (Li 2000). In the MTMD, the stiffness and the

damping coefficient were fixed while the mass of each of the TMDs were varied to

46

obtain variable frequency and damping ratio. As a comparison, the MTMD with fixed

mass and variable stiffness and damping coefficient is used and referred to as MTMD

(II). It was found that the optimum frequency spacing of the MTMD is the same as

that of the MTMD (II); the average damping ratio of the MTMD is a little larger than

that of the MTMD (II). In addition, it was demonstrated that the optimum MTMD is

more effective than the optimum MTMD (II) and a single optimum TMD.

2.5.1.4 Nonlinear Tuned Mass Dampers (NTMD)

It is well-known that the conventional TMD is sensitive to the structural variation

such as damage, mass variation or other sources. The TMD will lose its effectiveness

in reducing, or will even amplify, the structural responses when the frequency of the

structure shifts. To address the limitations of the TMD, research effort on nonlinear

Tuned Mass Damper (NTMD) with a nonlinear spring as illustrated in Figure 2-11.

Figure 2-11 Schematic model of nonlinear TMD (NTMD)

Roberson (Roberson 1952) first investigated an undamped nonlinear dynamic

vibration absorber consisting of a linear and a hardening cubic spring was investigate.

He defined a ‘suppression bandwidth’ as a frequency band between the response

peaks over which the normalized primary system response amplitude is less than

47

unity. It was demonstrated that the suppression band for the nonlinear TMD was much

wider than that of a TMD. This finding was later confirmed experimentally by Arnold

(Arnold 1955). Roberson’s work (Roberson 1952) was followed by Pipes (Pipes 1953)

using a hyperbolic sine spring without damping. The author concluded that the

nonlinearity in the spring can prevent the occurrence of sharp resonant peaks and to

introduce odd harmonic components of relatively small amplitude in the motion of the

absorber and primary system. In order to further improve the performance of the

NTMD, the Behaviour of a solid-type NTMD in reducing the primary structure

response was studied (Snowdon 1960). It was demonstrated that an NTMD with a

spring whose stiffness is proportional to frequency and a fixed damping factor can

reduce the resonance of the primary structure considerably.

Alternatives for the spring-dashpot NTMD, such as a triple-element NTMD, were

investigated and indicated that if a third element is introduced in series, a 15% to 30%

reduction can be obtained (Snowdon 1974). However, the promising reduction is quite

sensitive to the tuning of frequency. Meanwhile, Masri (Masri 1972) considered the

forced vibration of a family of piecewise linear two degrees of freedom (DOF)

dissipative non-autonomous systems. Exact solutions were obtained and the

asymptotic stability was confirmed. Analytical results showed that the properly

designed NTMD combing features of dynamic neutralizers, Lanchester dampers, and

impact dampers reduced some of the deficiencies inherent in the system, and was

better than the conventional forms of TMDs. Hunt and Nissen (Hunt et al. 1982)

introduced a viscous damper to an NTMD with a Belleville softening spring.

Softening force-displacement curves were illustrated where the softening nonlinearity

is controlled by the geometry of the Belleville washer. It was demonstrated that the

effective bandwidth could be doubled compared to the case of a conventional linear

48

TMD when a softening nonlinear spring is used. Use of such kinds of softening

nonlinear spring will greatly increase the possibilities of the NTMD to reduce the

unwanted vibrations to an acceptable level. Nissen et al. (Nissen et al. 1985), based on

the design procedure proposed in Hunt’s work (Hunt et al. 1982), realized the optimal

design for a softening NTMD with a Belleville spring from a technical perspective,

aiming to maximize the effective bandwidth. Afterwards, Jordanov et al. (Jordanov et

al. 1989) proposed a numerical method for optimal design for linear and a nonlinear

TMD in undamped and damped primary systems. The method of sounding was

utilized to examine the objective functions and to find the optimal solution under

multi-criteria. The effectiveness of the proposed algorithm for searching the optimal

design was demonstrated in the research.

Natsiavas (Natsiavas 1992) investigated the steady-state solutions and the stability

characteristics of a nonlinear system consisting of a nonlinear primary structure and a

weakly nonlinear TMD. An averaging method was used to obtain the approximate

steady-state solution and Eigen analysis was performed to characterize the stability of

the located oscillations. Two types of stability were encountered: one is the

saddlenode type bifurcation and the other is the Hopf bifurcation. Based on the

obtained steady-state solution, a parametric study was implemented for three cases:

linear primary structure plus nonlinear TMD, nonlinear primary structure with linear

TMD, and nonlinear primary structure with nonlinear TMD. In each case, the

representative result was illustrated. It was indicated that the selection of proper

parameters of the nonlinear TMD would result in substantial improvements and avoid

potentially dangerous effects.

49

2.5.1.5 Pendulum Tuned Mass Damper (PTMD)

In addition to the conventional TMDs, the pendulum TMDs (PTMD) consisting of

a cable and a mass suspended at the top part of a building has received popularity in

the community of vibration control in recent years. The most notable and largest TMD

ever constructed in real world is the pendulum TMD installed on top part of Taipei

101 building as illustrated in Figure 2-12.

Figure 2-12 Illustration of the PTMD installed in Taipei 101 (Soong and Spencer 2002)

Nagaragaiah (Nagaragaiah 2009) used the PTMD to control excessive floor

vibrations due to human activities. His results indicated that a properly-tuned PTMD

can effectively control the floor vibrations while an off-tuned PTMD may not function

effectively. In order to overcome the off-tuning of PTMD, Nagaragaiah (Nagaragaiah

2009) further proposed a concept of adaptive-length pendulum TMD (APL-PTMD). It

was shown, experimentally, that the APL-PTMD can significantly reduce the

structural responses and outperforms its equivalent passive counterpart. Nagarajaiah,

50

(Nagarajaiah 2009) the author, also proposed the idea of using a rolling ball on a

controllable guiding surface which is conceptually equivalent to a pendulum TMD but

is easier to vary the radius. This idea was then confirmed by Matta et al. (Matta et al.

2009) where a rolling ball pendulum TMD moving on a three-dimensional guiding

surface was studied theoretically and experimentally. The authors showed that this

rolling ball pendulum TMD can reduce the structural responses in two mutually

orthogonal horizontal directions.

Sun et al. (Sun et al. under review) further analysed the performance of the

STMD/APTMD under harmonic excitation and ground motion where closed-form

solutions were derived. Their results provided physical interpretation with respect to

how the design parameters influence the reduction effect of the STMD/APTMD.

Based on the results presented by Sun et al. (Sun et al. under review) an optimal

design for the STMD was proposed. In order to experimentally validate the results

presented by Sun et al. (Sun et al. 2013), they used an adaptive pendulum TMD

(APTMD) and an NTMD in parallel to attenuate the response of a Duffing system.

Their results indicated that when an NTMD is used alone, a high amplitude detached

resonance in the lower frequency range is identified. When the APTMD is used, the

high amplitude detached resonance is greatly attenuated and significant attenuation of

the structural responses over a large frequency range can be obtained. In addition, the

APTMD can prevent the occurrence of the “jump phenomenon” existing in a

nonlinear dynamic system. Of course, nonlinearity will be involved when the

pendulum TMD experiences large displacement. As a matter of fact, a pendulum

TMD is essentially a nonlinear TMD with softening nonlinearity (Bajaj 1994). In

other words, the frequency response curve of a pendulum TMD, when large rotation

happens, leans to the left. In addition, those nonlinear characteristics, such as

51

bifurcation and chaos within a nonlinear dynamic system can also result when a

pendulum TMD is used. Research efforts on this aspect can be found in Lee et al. (Lee

et al. 1999) and Song et al. (Song et al. 2003).

2.5.1.6 Base Isolation

Through a proper initial design or subsequent modifications, building or non-building

structures could survive the impact of a potentially devastating seismic impact (Kitagawa

and Midorikawa 1998). In these systems, a flexible isolation system is introduced between

the foundation and superstructure to increase the natural period of the system. Increment

in flexibility typically results in the deflection of a major portion of the earthquake

energy, reducing accelerations in the superstructure, while increasing the displacement

across the isolation level.

Figure 2-13: Difference between lateral deformation in controlled and uncontrolled

systems (Otani 1981)

It can be applied both to a newly-designed building and to seismic upgrading of the

existing structures (Zhou and Xian 2001). Normally, excavations are made around the

building and the building is separated from the foundations (Soong and Spencer

52

2002). Steel or reinforced concrete beams replace the connections to the foundations

by the isolating pads or base isolators. The base isolation tends to restrict transmission

of the ground motion to the building; it also keeps the buildings' position properly

over the foundation (Seki, Vacareanu et al. 2008). By their response to an earthquake

impact, all isolation units may be divided into two basic groups: shear units and

sliding units. This technology can be used both for new structural design and seismic

retrofitting (Jangid and Kelly 2001). The difference between roof displacement of

controlled and uncontrolled systems and the shape of foundation of a structure with

base isolation controller are shown in Figure 2-13 (Otani 1981)

2.5.1.7 Viscous Fluid Dampers (VF)

Viscous Fluid damper operates based on the principle of fluid flow through orifices.

They dissipate energy through the movement of a piston in a highly VF based on the

concept of fluid orificing, i.e. a stainless steel piston moves through the chambers that are

filled with inert, non-flammable, and non-toxic silicone oil that remains stable for over

long periods of time (Taylor and Constantinou 1998). The pressure difference between the

two chambers causes silicone oil to flow through an orifice in the piston head and seismic

energy is transformed into heat, which dissipates into the atmosphere (Taylor and Duflot

2005). These devices are attractive for incorporation in diagonal bracing systems. By

requiring no external power source and little maintenance, they have become very

attractive for structural applications. Viscous fluid dampers used in a steel bridge structure

are shown in Figure 2-14.

53

Figure 2-14: Typical viscous fluid dampers used in diagonal bracings (Hemalatha and Jaya 2008)

Viscous Fluid dampers typically exhibit very high energy dissipation capacity in

comparison to their physical size (Jagadish, Prasad et al. 1979, Kitagawa and Midorikawa

1998). This form of damper dissipates energy by applying a withstanding force over a

finite displacement through the action of a piston forced through a fluid-filled chamber for

a completely viscous, linear behaviour, or in a damping wall which uses a full-storey steel

plate traveling in a wall filled with viscous materials to provide added damping (Carlson

and Jolly 2000).

2.5.1.8 Viscoelastic Dampers (VE)

Viscoelastic materials used in structural applications are usually copolymers or glassy

substances that dissipate energy through shear deformation. Therefore, energy dissipation

takes place when relative movements occur between the centre plate and the outer steel

flanges. In order to incorporate the mechanical effects from VE dampers into the

structural dynamic design, it is important to use a proper force-deformation model to

correctly describe the frequency dependence of the damper (Fu and Kasai 1998). A

54

typical VE damper, which consists of VE layers sandwiched between steel plates is

shown in Figure 2-15. Therefore, energy dissipation takes place when relative

movements occur between the central plate and the outer steel flanges.

Figure 2-15: Typical VE Damper Configuration (Soong and Spencer 2002)

2.5.2 Semi-Active Controllers

A semi-active control system may be defined as a system which typically requires a

small external power source for operation (e.g. Battery) and utilizes motion of the

structure to develop control forces. The magnitude of control force can be adjusted by

the external power source (Aldemir 2003). Control forces are improved based on the

feedback from the sensors that measure the excitation and/or response of the structure.

In general, the performance of the structure with the semi-active control system is

superior to the one with a passive control system, while simultaneously requiring

smaller control forces (Ji, Moon et al. 2005). A semi-active control system is generally

originated from a passive control system which has been subsequently modified to

allow for the adjustment of mechanical properties. Control forces are developed

through an appropriate (based on a pre-determined control algorithm) adjustment of

the mechanical properties of the semi-active control system. Furthermore, control

forces in many semi-active control systems initially act to oppose the motion of the

55

structural system and therefore increase global stability of the structure (Zhou and

Xian 2001). The role of semi-active controllers in a structure is shown in Figure 2-16.

One challenge in the use of semi-active technology is in developing nonlinear control

algorithms that are appropriate for implementation in full-scale structures (Symans

1999).

Figure 2-16: Structure with a Semi-Active Control System (Symans 1999)

2.5.3 Active Control of Structures

An active control system is defined as a system which typically requires an external

power source for the performance of electro-hydraulic or electro-mechanical actuators

which supply large control forces to the structure. Control forces are developed based on

the feedback from the sensors that evaluate the excitation and/or response of the structure

(Datta 2003). Active structural control system consists of three parts as follows:

1. Sensors located inside the building to measure external excitations and/or

structural response variables;

2. Devices to process the measured information from the sensors and to compute

necessary control forces needed based on a given control algorithm; and

3. Actuators, usually powered by external sources, to produce the required forces.

56

Figure 2-17: Structure with Active Control System (Symans 1999)

The role of active controllers in a structure is shown in Figure 2-17. The recorded

measurements from the response and/or excitation are monitored by a controller (a

computer) based on a pre-determined control algorithm that determines the appropriate

control signal for the operation of actuators (Agrawal, Fujino et al. 1993). The generation

of control forces by electro-hydraulic actuators requires significant power, which starts

from tens of kilowatts for small structures to megawatts for large structures. The initial

effect of some experimentally tested active control systems was to adjust the level of

damping with a minor modification of stiffness (Yang and Soong 1988). Advantages

associated with active control systems are cited below:

1. Enhanced effectiveness in response control: Degree of effectiveness is only limited

by the capacity of the control systems.

2. Relative insensitivity to site conditions and ground motion.

3. Feasibility to multi-hazard mitigation situations: An active system can be used, for

example, for motion control against both strong wind and earthquakes.

4. Selectivity of control objectives: One may be more important than others; for

example, human comfort is more important than other aspects of structural motion during

57

noncritical times, whereas increased structural safety may be the objective during severe

dynamic loading (Soong and Spencer 2002).

Three major classes of control systems described above are sometimes combined to

form a new kind of controller called hybrid control systems (Tabuada 2009). Hybrid

control systems consist of combined passive and active devices or combined passive and

semi-active devices. The role of hybrid controllers in a structure is shown in Figure 2-18

(Yang and Agrawal 2002).

Figure 2-18: Structure with a Hybrid Control System (Symans 1999)

2.6 Analytical Method for Analysing Nonlinear Systems

Because a nonlinear dynamic system is very difficult to analyse directly,

researchers resorted to seek the approximate solutions. Perturbation methods, which

are widely used in the field of nonlinear dynamics, are a technique in which an

approximate solution is achieved in an asymptotic fashion. This approach can produce

accurate results for weakly nonlinear structure experiencing relatively small

oscillations. However, its accuracy will be weakened when strong nonlinearity and

large displacement is experienced by the system.

58

2.6.1 Perturbation Method: Multiple Scales Method

The perturbation method is always used to achieve the periodic solution of a

nonlinear dynamic system. Several efficient methods that are frequently used include

the Av eraging Method, the Harmonic Balance Method, the Lindstedt-Poincare

Method, the Multiple Scales Method and so forth (Nayfeh 1973). The procedure of

computing the approximate solution using the Multiple Scales method is briefly

introduced here. A general nonlinear dynamic system can be represented by:

(2-12)

Where x denotes the state-space vector; M is an m-dimensional parameter vector.

Showing the time scales using a small parameter ε, i.e.

Tn=εnt (2-13)

Then the derivative with respect to time t can be represented by these slower time

scales as:

· (2-14)

Assuming the solution x(Hemalatha)can be expanded to a series:

(2-15)

Substituting Equation (2-15) and Equation (2-14) into Equation (2-12) yields:

(2-16)

0 = M); x F(x,

101

1

0

0 ...dt

d DDTdt

dTTdt

dT

...,...,,,...,,,...,,, 21022

21012100 TTTxTTTxTTTxtx

0,...,,,,,,...,,, 32132

321 DDDxxxF

59

Since the small parameter ε is arbitrary, the coefficients of the ε with different

orders should equal zero to make the equation hold, i.e. Then a set of equations are

obtained. Solving these equations from the low order term to high order term yields

the expression for x1, x2, x3….Eventually, the frequency response function, which is a

nonlinear algebraic equation, can be obtained by means of eliminating the scalar

terms. Then the frequency response curve can be calculated through numerically

solving the nonlinear algebraic equation. Because the solution of the frequency

response function is multi-valued for which the stability needs to be determined. This

is discussed in the following subsection.

2.6.2 Local Stability Analysis

Stability of the solution needs to be determined through eigen analysis. The general

Equation (2-12) can be written as:

(2-17)

Let X0(t) be the solution of the system at parameter M0, i.e. X0(t) = G(t;M0). Adding

a small disturbance y(t) to X0(t), i.e.

x(t) = X0(t)+y(t) (2-18)

Substituting Equation (2-18) into Equation (2-17), expanding the result and

retaining the linear terms in disturbance produces:

(2-19)

Equation (2-19) reduces to:

M)xG(t; x

2

000 )();(

)MG(t; M)y(t); (t)G(X (t)y (t)X yOtyxMtG

60

(2-20)

Local stability of the solution X0 can be determined be means of analyzing the eigen

values of the matrix A. If A is a constant matrix, its eigen value and eigen vector can

be calculated directly. Otherwise, Floquet theory is needed to determine the local

stability also by means of analyzing the eigen value of a monodromy matrix (Nayfeh

1995). If the real part of all the eigen values are negative, the disturbance will

eventually vanish and the examined solution is asymptotically stable. On the other

hand, the solution is unstable if the real part of the eigen values are positive. In the

case that the real part of the eigen value is zero, bifurcations will occur, which can be

analyzed by keeping higher order terms when Taylor expanding Equation (2-18). In

addition, if the real eigen value changes sign, a Saddle-node bifurcation might occur

and if a pair of complex conjugate eigenvalues whose real part changes sign, the Hopf

bifurcation resulting in quasi-periodic oscillations might occur. It is noted that the

rules listed here are general description for determining the local stability of the

solution.

2.7 Numerical Methods for Analysing Nonlinear Systems

In comparison with the analytical methods, numerical approaches are suitable for

solving problems involving strong nonlinearity and large displacement. With the rapid

progress of computers and high-performance computing (HPC), numerical

computation is playing a more and more important role in the field of science and

technology. As for computing the solutions for nonlinear systems, two methods are

always used: one is the Time Integration Method and the other is the parameter

Continuation Method. The following subsections will introduce the principles of each

of the two methods.

)yMA(t; (t)y 0

61

2.7.1 Time Integration Method

The most direct numerical method for computing the periodic solutions of a

dynamic system is the so-called brute-force approach (Nayfeh 1995) which is

essentially based on time integration. In this approach, the system is integrated at a

given initial condition for a long enough time until the steady-state solution is reached.

The advantage of this approach is that it is clear and easy to implement. Meanwhile, it

is very general because it can calculate fixed points, periodic solutions, quasi-periodic

solutions and chaotic solutions. Mathematically, a dynamic system can be represented

by a set of ordinary differential equations (ODEs) or partial differential equations

(PDEs) together with the initial values (IV) and the boundary values (BV). Problems

with initial values are referred to as IVP and the latter is BVP.

Approximate numerical solutions for the IVP can be obtained through time

integration. Generally, the numerical methods for IVP can fall into two large

categories: explicit method and implicit method. Explicit method, as the name

suggests, the value of the variables at the right hand side (R.H.S.) of an equation at

each time step are known; hence, no iterations are needed to compute the value for the

next step. In comparison, there are unknown variables at the R.H.S. when computing

the value for the next time step and iterations are performed until the convergence

criterion is satisfied. Generally, the implicit method has higher order accuracy and is

more stable than the explicit method. Forward Euler method, the Runge-kutta method,

and the Gaussian quadrature method belong to explicit methods. Implicit methods

include Back Euler method, the Adams-Moulton method (JohnC 2003), the well-

known Newmark-β method which is widely used in structural engineering and so

forth.

62

2.7.2 Continuation Method

Although the time-integration method is simple and general, it has several

disadvantages:

(1) for lightly damped system, the convergence can be very slow because the

transient response takes a long time to decay, (2) not all the unstable solutions can be

located by reversing the direction of integration and, (3) it is difficult to judge whether

the steady-state solution is achieved. In order to overcome these drawbacks, a more

direct approach is proposed and used here.

In comparison with the time integration method, Continuation method for periodic

solutions computes the solution through generating a continuum of periodic solutions

with respect to a control parameter, say α, rather than direct time integration. The

solution seeking procedure starts from an initial guess, say x0, which is the solution

corresponding to the starting point of the continuation parameter, say α0. The initial

guess can be obtained either analytically or numerically. Then the continuation

parameter α is varied from α0 to α1 and the initial solution x0 is updated to x1 by

means of solving a set of algebraic equations resulting from discretizing the original

dynamic system. The procedure is similar to that for solving boundary value problems

(BVPs). Arc length method, which is widely used in BVPs, is used in some softwares,

like DERPER (Holodniok 1984). Another similar method, Pseudo-Archlength method

(Doedel 1986, 1991) which uses a pseudo-arc length constraint equation (arc length

constraint equation is used in the Arc length continuation method), is used in AUTO

(Doedel 1987) bifurcation and continuation software. The principle behind the Arc

length method and the Pseudo-Arc length method are illustrated here:

63

Arc length Continuation

For a dynamic system defined by Equation (2-12), let x(T(s), )(s), )(s)) with period

T(s) be a periodic solution of the equation, where the arc length s is used as the

continuation parameter and η denotes the state-space variable. Hence, the solution

x(T(s), n(s), n(s)) satisfies the following equation:

G(T(s,f), the following equations,− η(s) = 0 (2-21)

with the initial condition: x(T(0), )(0), )(0)) = η(0). Differentiating Equation (2-

21) with respect to s yields:

(2-22)

where (˙) denotes derivative with respect to s; ∂ den is a n×n matrix, ∂atri is a

n×1 matrix, ∂ maT is a n×1 matrix. Equation (2-22) contains n linear algebraic

equations while the system has (n+ 2) unknowns . Therefore, two additional

equations are needed. One additional equation comes from the Euclidean arc length

normalization:

(2-23)

Another equation is specified in the form of a phase condition, i.e. one variable ηk

in the vector η along the continuation path is fixed.

(2.24)

0,,G

,,G

,,T

G

TTTT

,,T

122 TT

0s

k

dd

64

With the two additional equations, the (n+2) equations can be solved for the

periodic solution x(T, equ).

Pseudo-Arc length Continuation

The principle behind Pseudo-Arc length Continuation method differs from the Arc

length method in that the two additional equations are specified in a similar but

different manner. The first equation is specified using a phase condition in an integral

form. Let s0 and s be the two consecutive points on the branch; x0 = x(T0, 0, 0),

are the two periodic solutions corresponding to the two points s0 and s. If is a

solution, then is also a solution for any σ. Then the phase condition is

obtained when the distance is minimized with respect to the time variation :

(2-25)

Setting dD(σ)/dd = 0 produces.

(2-26)

Assuming the solution of Eq. (3.15) is σ∗, i.e. (3.14) reduces to:

(2.27)

Integrating Equation (2-27) by parts and using Equation (2-12) produces:

(t)x

x

) +(t x

x-x 0

dtDT 2

0

0 (t)x-)(tx

dtd

T 2

0

0 (t)x-)(tx

) +(t x

dtxxdtxxT

xxdtt TTTTT

00

0

0 02

1(t)x-x(t)

65

(2-28)

The second equation is specified by the pseudo-arc length constraint:

(2-29)

where (˙) designates derivative with respect to the arc length s and δs represents

the step size along the continuation path. Equation (2-22) together with Equation (2-

28) and Equation (2-29) constitute the (n+2) equations for the dynamic system with (n

+ 2) unknowns. Then the solution x(T(s),η(s),α(s)) can be achieved by means of

solving the (n + 2) linear algebraic equations.

2.8 Summary

General information about earthquake and seismic activity, earthquake-resistant

design for structural buildings and review of AS1170.4, Australian Standard, have

been reviewed at the beginning of this chapter. Additionally, an overview with respect

to the research effort focusing on several kinds of those widely used TMDs, including

the conventional passive TMDs, the LTMDs, the NTMDs, the MTMDs and the

STMDs, and the related variable damping and stiffness devices are presented in this

chapter. The TMDs, which have been well understood and widely deployed in real

engineering applications, have their limitations due to the narrow effective

suppression bandwidth. In comparison, The MTMDs and the NTMDs can effectively

broaden the suppression bandwidth. However, the application of NTMDs is more

convenient than that of the MTMDs, which needs a lot of effort in the process of

design and installation. At the same time, it is demonstrated that the STMDs can

dtTxFxxdtxtx TT

TTT

0000

0

00

0

;,,

s

T

TTTdtxxtx 00000

0

0)(

66

provide comparable response reduction to that of the active TMDs yet they require an

order of magnitude less power. Therefore, the NTMDs and the STMDs are the focus

of this thesis. In addition, analytical and numerical methods used to solve nonlinear

dynamic equations have been reviewed in this chapter. To sum up, the Continuation

Method can be used to efficiently and accurately trace stable and unstable solution

branches with respect to a predetermined control parameter. This method is especially

useful in the analysis of nonlinear systems due to the complex and often unpredictable

response behaviour. Additionally, the Psuedo-Arc length method provides the unique

ability to trace folding solution branches. Therefore, in the current thesis, the

bifurcation continuation software AUTO (Doedel 1997), which is based on the

Psuedo-Arc length method, is used to compute the responses of the nonlinear dynamic

systems.

67

CHAPTER 3

LITERATURE REVIEW OF

FAÇADE SYSTEMS

68

3.1 Introduction

A façade is generally one exterior side of a building, usually but not always, the front.

The word comes from French, literally meaning "frontage" or "face". The aluminium

frame which consists of mullions and transoms is normally in-filled with glass that

provides an architecturally pleasing skin as well as advantages such as natural day

lighting. From the architectural viewpoint, facade of a building is very important from

the design standpoint, since it sets the character for the rest of the building. Façade also

provides shielding against environmental factors like wind or rain and provides light and

ventilation to the structure (G. James Glass and Aluminium Pty Ltd 2003, Hareer 2007)

Figure 3-1: Typical components of a façade panel (Olanders Window Replacement 2011)

Fast developments during the 19th century era, when industrial revolution happened,

led to major advances in structural technologies. In the field of facade systems, these

advances resulted in the usage of industrialized components in the installation. Also, size

of facade components and their strength and durability have been improved. After

significant changes in the field of structural design, the role of façade system has become

more noteworthy to be considered. Two famous materials which have been used widely

since 1930 are precast concrete and aluminium (Zhang, Provis et al. 2014). After World

69

War II, when usage of façade in buildings came to a temporary halt, rapid development

which happened in building materials opened up a new view of the façade. Construction

of facade significantly increased and reached an incredible boom during the mid to late

60s (Priwer and Phillips 2014). Facade panel is steadily an expensive part of a building

construction which amounts to about 20% of total building costs (Eicker and Pietruschka

2009). Special attention should be paid to its protection from damage or collapse. In

modern skyscrapers, exterior walls are often suspended from the concrete floor slabs

(Murray 2009). Examples include curtain walls and precast concrete walls. In general,

the facade systems that are suspended or attached to the precast concrete slabs will be

made from aluminium (powder coated or anodized) or stainless steel (Brookes and Grech

2013). Typical glazing panels consist of the elements shown in detail in Figure 3-1:

Frame: aluminium frame typically consists of horizontal components, which are

called transom, and vertical components called mullions.

Glass: air or gas fills between two panes of glass space. Special Low-E coating on

the glass blocks infrared light to keep heat inside in the winter and outside in the

summer. It also filters damaging ultraviolet light (UV) to help protect interior

furnishings from fading.

Spacer: a spacer keeps a window's dual glass panes at correct distance apart for

optimal airflow between the panes. Too much or too little airflow can affect the

efficiency of insulating glass. The design and material of the spacer can also make

a big difference in the ability to handle expansion and contraction and thus reduce

condensation. Insulating spacers between the panes of glass reduce heat transfer

and condensation (Olanders Window Replacement 2011).

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3.2 Types of Facade Systems

More than 70 different kinds of facade systems have been defined in terms of

shape, weight, and performance by architectural designers. But, generally speaking,

panels are categorized as heavy cladding, light weight cladding, and in-fills, as shown

in Table 3-1. There is also the possibility of a combination of the two called a mixed

system. Cladding is attached externally to the primary structure, whereas in-fills are

constructed within the frame of the structure.

Table 3-1: Different kinds of façade systems

Infills Heavyweight Cladding Lightweight Cladding Glazing Infills Precast concrete panels Curtain wall Masonry Infills Stone panels Stick curtain

- - Spider glazing - - Brick veneer - - Double skin

3.2.1 Infills

Clay bricks or concrete masonry blocks are heavy rigid materials that have been

conventionally used as in-fill units. This type of system is simple to construct and is

particularly prevalent in low- to medium-rise office structures. Masonry in-fill

constructions have been built in medium-risk seismic regions, but not high seismic

regions for several decades because of concerns about their poor seismic performance

and complexity of their interaction with structures (AAMA 501.4 2000). However,

more light weight in-fill panels such as light steel- or timber-framed in-fill walls (dry

walls) are now available (Tasligedik, Pampanin et al. 2012) . In-fill panels are often

combined with a glazing in-fill system consisting of an aluminium frame attached

directly to the in-fill panel or structure (Rice 2006). Some minor in-plane movement

of panels is controlled by rubber gaskets which hold the panes of glass in place (Baird,

71

Palermo et al. 2012). These systems are normally located within the frame of the

structure.

3.2.2 Light Weight Cladding

Lightweight cladding incorporates large sized glazing which in turn includes a

broad range of facade systems where each typology of lightweight cladding can also

include a wide range of systems.

3.2.2.1 Stick System

Stick curtains are very common and versatile and can be used for any kind of

building from glass high-rise to single storey shop fronts (Permasteelisa 2009).

Because of the number of joints in stick curtain walling, it has generally been very

good in accommodating variability and movement in the building frame. It is also

suitable for irregularly-shaped buildings.

Figure 3-2: Stick system façade (Permasteelisa 2009)

Assembly is slower than pre-assembled systems (Walker III, Niemoeller et al.

2011). A typical stick system and its installation are shown in Figure 3-2 and Figure

72

3-3 (Permasteelisa 2009). A stick system consists of a framework of horizontal and

vertical framing members. Into the framework, the in-fill units are fitted and may

constitute a mixture of fixed and opening glazing and insulated panels. The elements

are prepared at the plant and, afterwards, assembled on site as a kit of parts. The

mullions are typically spaced between 1.0 and 1.8 m. The glazed or opaque panel is

retained with a pressure plate or clamping element and screw-fixed every 150 to 300

mm. Sometimes, hammer-in structural gaskets are used instead. The pressure plate is

mostly covered with a snap-on decorative element (Patterson 2011).

Figure 3-3: Typical assembly of stick system façade (Permasteelisa 2009)

3.2.2.2 Curtain Wall

Curtain wall is a kind of barrier which separates the exterior of a building from its

interior. It plays a vital role in the aesthetic appeal of the primary building and has the

following crucial roles (He 2005):

Wind/rain/water protection

73

Insulation against hot and cold climates

Protection from noise and pollution

The curtain wall facade does not carry any dead-load from the building other than

its own weight. The wall transfers horizontal wind loads acting upon it to the main

building structure through connections at floors or columns of the building. Curtain

walls are designed to span multiple floors and consider design requirements such as

thermal expansion and contraction, building sway and movement, water diversion and

thermal efficiency for cost-effective heating, and cooling and lighting in the building

(Permasteelisa 2009). Framed-glazed curtain walls are typically designed with

extruded aluminium members although early curtain walls were made of steel. They

are designed to resist air and water infiltration, sway induced by wind and seismic

forces acting on the building, and its own weight. Curtain walls can be divided into

these groups (Kragh and Components 2001):

3.2.2.3 Unitized Curtain Wall

Unitized curtain walls are pre-fabricated; so, mechanical handling is required to

position, align, and fix units onto pre-positioned brackets attached to the concrete

floor slab or structural frame (Speck 2010). They span floor to floor and are anchored

to the building's load-bearing structure. Pre-fabrication of this type of facade allows

for better quality controls, makes installation very quick, does not require the use of

scaffolding, and minimizes work in the worksite with lower installation costs. Typical

unitized curtain walls and their installation are shown in Figure 3-4 and Figure 3-5.

74

Figure 3-4: Unitized Curtain Wall (Permasteelisa 2009)

The system is more complex in terms of frame design in comparison to stick system

and the possibility of creating complex and/or irregular surfaces is limited. They have

higher direct costs and are less common than stick systems (Chew, Tan et al. 2004).

Fewer site staff is needed in comparison with stick systems, which can make the

systems more cost effective.

Figure 3-5: Installation of curtain wall (Permasteelisa 2009)

75

3.2.2.4 Panelized Curtain Wall

Panelised curtain walling is composed of large prefabricated panels of bay width

and storey height, which are connected back to the primary structural columns or to

the floor slabs close to the primary structure (Lindow and Jasinski 2003). Fixing the

panels close to the columns reduces problems due to the slab deflection in the mid

span which affects stick and unitized systems (Webb 1989). Panels may be of precast

concrete or comprise a structural steel framework, which can be used to support most

of the cladding materials (e.g. stone, metal, and masonry). A typical panelized curtain

wall is shown in Figure 3-6.

Figure 3-6: Panelized curtain wall (Permasteelisa 2009)

3.2.2.5 Spandrel Panel Ribbon Glazing

Spandrel panel ribbon glazing is a long or continuous run of vision units fixed

between spandrel panels supported by vertical columns or the floor slabs (Hinman and

Arnold 2010). They can be used in conjunction with spandrel panels, i.e. horizontally

spanning prefabricated or precast concrete units. It may also be used with spandrels

composed of stand walls which are faced with rain screen panels (Mazzoni, Bowser et

76

al. 1976). Typical spandrel panel ribbon glazing system in a multi storey structure is

shown in Figure 3-7.

Figure 3-7: Example of spandrel panel ribbon glazing (Permasteelisa 2009)

3.2.2.6 Bolted Glass Façade

This type of facade is created to fulfil an architectural and functional requirement

for maximum transparency. It eliminates all opaque supporting elements and does not

employ sticks (Vyzantiadou and Avdelas 2004). Glazed panels are suspended using

the lightest possible systems available. There are two different types of glass facades:

the independent assembly and suspended assembly, shown in Figure 3-8 and Figure

3-9 (Carmody, Selkowitz et al. 2004).

Figure 3-8: Independent assembly (Permasteelisa 2009)

77

Figure 3-9: Suspended assembly (Permasteelisa 2009)

3.2.2.7 Double Skin Façade

3.2.2.7.1 Definition

Designing and building energy-efficient residential and commercial buildings is a

priority for building and construction industry. Attention of government

administrators and building owners to environmentally-friendly structures has been

drawn to these state-of-the-art concepts. It is also expected to move to higher star

rating buildings that deliver better energy and thermal efficiency and comfort (Arons

2000). Developments of a new kind of facade systems have been boosted because of

energy performance concerns of previous facade technologies (da Silva and Gomes

2008). Transparency and visual attraction are other crucial factors to be considered in

high glass skyscrapers. Based on those essential elements, double-skin or multi-skin

facade (also known as active envelope) systems are recently presented as a valuable

solution for achieving the aforementioned goals in modern architecture (Poizaris

2004). The double‐skin facade systems are the architectural concept driven by

aesthetics and desire for mostly all‐glass high-rise buildings. They consist of two

panes separated by a cavity through which air can circulate naturally or mechanically.

78

A typical double-skin facade system is shown in Figure 3-10 (Poizaris 2004) . In most

cases, a shading device is provided in the cavity (Hensen, Bartak et al. 2002). Higher

prices due to complexity of design and installation can be justified by their increasing

demands because of wide transparent surfaces and high thermal performance

(Streicher, Heimrath et al. 2007).

Figure 3-10: Typical Double-Skin-Façade System (Poizaris 2004)

For evaluating the installation of double skin facades as a glazing envelope of

buildings, factors of climate, orientation, detailing, and construction cost and energy

price should be considered. They should be evaluated in terms of their relative

advantages in relation to those factors (Poirazis 2008). Double-skin facades can be

described as a traditional single-facade doubled inside or outside by a second and

essentially glazed facade. Apart from the type of the ventilation inside the cavity, the

origin and destination of the air can differ depending mostly on climatic conditions,

usage, location, occupational hours of building, and the HVAC strategy. Each of these

two façade layers is commonly called a skin and these skins are placed in such a way

that air flows in the intermediate cavity (Poirazis 2008). Through the cavity, for

example, hot air can be effectively removed in summer time and also natural

79

ventilation can be introduced even at higher levels of tall buildings, because there are

additional exterior skins which act as wind buffers.

Figure 3-11: Exterior Circulation Double Skin Curtain Wall (Arons 2000)

The glass skins can be single- or double-glazing units and are often for protection

and heat extraction reasons during the cooling period, with solar shading devices

positioned inside the cavity (D. Saelens 2003).

Figure 3-12: Facade detail: Hot expelled at each floor, cool air drawn in (Lee, Selkowitz

et al. 2002)

In Figure 3-11 and Figure 3-12 the way ventilation works in the system is shown

(Arons 2000). The DSF system reduces energy usage in buildings which potentially

80

results in economic benefits in the long run, even though their initial construction cost

is higher than that of conventional single-skin facades (Shameri, Alghoul et al. 2011).

DSFs have the potential to reduce building heating and cooling energy consumption in

several ways; however, not all DSFs built in recent years perform well (Gratia and De

Herde 2007). Furthermore, in most cases, large air-conditioning systems have to make

up for summer overheating problems and energy consumption often exceeds the

intended heating energy savings. Other concerns about DSF performance include fire

safety (fires spreading between floors via the cavity) and their maintenance is costly

(Zhou and Chen 2010).

3.2.2.7.2 History

Jean-Baptiste Jobard, Director of Industrial Museum in Brussels, described an early

version of a mechanically ventilated multiple-skin facade in 1849. He mentioned how

in winter, hot air should be circulated between two glazing, while, in summer, it

should be cold air (Saelens, Roels et al. 2003). The first instance of a double-skin

curtain wall appeared in 1903 in Steiff Factory in Giengen/ Brenz. Priorities were to

maximize day lighting while taking into account the cold weather and strong winds of

the region (Saelens, Blocken et al. 2005). The solution was a three-storey structure

with a ground floor for storage space and two upper floors for work areas. The

structure of the building proved to be successful and two additions were built in 1904

and 1908 with the same double-skin system, but using timber, instead of steel, in the

structure for economic reasons. The building is shown in Figure 3-13 (Streicher,

Heimrath et al. 2007).

81

Figure 3-13: Steiff factory, Giengen/Brenz, Germany (Streicher, Heimrath et al. 2007).

In Russia, Moisei Ginzburg ran an experiment with double-skin strips in the

communal housing blocks of his Narkomfin building (1928) and Le Corbusier

designed Centrosoyuz in Moscow. A year later, Le Corbusier started the design for the

Cite de Refuge (1929) and the Immeuble Clarte (1930) in Paris and postulated two

new features. Little or no progress was made in double-skin glass construction until

the late 1970s and early 1980s. During the 1980s, this type of facade started gaining

momentum. Most of them were designed while taking environmental concerns, like

offices of Leslie and Godwin, into account.

In other cases, the aesthetic effect of multiple layers of glass was the principal

concern. In the 1990s, two factors strongly influenced the proliferation of DSFs.

Environmental concerns started influencing architectural design both from a technical

standpoint and as a political influence that made "green buildings" a good image for

corporate architecture (Braham 2005)

3.2.2.7.3 Examples

Examples of notable buildings which utilize a double skin facade are 30 St Mary

Axe (also known as The Gherkin) and 1 Angel Square both in the UK (Allinson

82

2007). Both of these buildings achieved great environmental credentials for their size

with the benefits of a double-skin key. The Gherkin features triangular windows on

the outer skin which skelter up the skyscraper. These windows are opened according

to weather and building data and allow more or less air to flow through the building

for ventilation (Roth 2007).

3.2.3 Heavyweight Cladding

Heavy claddings can be defined as having a mass of more than 80 kg/m2 (Baird,

Palermo et al. 2011). Most of the heavy claddings surveyed during previous

earthquakes are precast concrete panel systems; so, the window system inside the

panels could have been classified as a glass in-fill, because the surrounding panels

have such high in-plane stiffness that no allowance for movement is made for these

window systems.

Figure 3-14: Precast facade panel installations (Traulsen and McClellan 2010)

83

Precast concrete panels have been the most popular cladding material used in new

non-residential buildings over the past decade (Palermo, Pampanin et al. 2010). They

can be either storey-height panels that provide multiple architectural functions, or

panels that are purely aesthetic. Typical precast façade panels and their installation are

shown in Figure 3-14 (Traulsen and McClellan 2010).

3.3 Typical Facade Connections and their Inherent Problems

Cladding connections can have numerous configurations; however, they are

typically attached to either the beams or columns of primary structures. Insufficient

connection strength and low damping capability of facade components are reported as

the main cause of failure in most buildings (Palermo, Pampanin et al. 2010). These

elements may be susceptible to excessive inter-storey deformations and noticeable

accelerations. Glass panel’s damage can be a threat to pedestrians and the cause of

additional cost for repair or reinstallation.

Throughout an earthquake excitation, the behaviour of the facade panels is

governed by the cyclic interaction between the panels and the supporting primary

structure and usually three primary effects are applied simultaneously to the facade

and connections. (1) Accelerations of the panel generate inertia forces which are

transmitted from the panel to the main structure via shear loading of the connectors

(Baird, Palermo et al. 2012). (2) Horizontal inter-storey drift resisted by the panels,

which results in horizontal shear forces in the connection, and (3) panel’s gravity load

which is carried by the bearing connections. The success of facade systems is related

to the capability of the facade connections in meeting both strength and ductility

requirements (Hunt 2010).

84

3.3.1 Bearing Connection

For the bearing connections, the connector elements in both horizontal and vertical

directions are assumed to be very stiff and very strong in order to transfer the self-

weight of the panel to the structure. The fixed bearing connections support the gravity

loads of the cladding, whereas light weight claddings are generally fixed to the

structure with connections that do not allow movement; therefore, they must be able to

tolerate relative displacement within the system (Palermo, Pampanin et al. 2010).

3.3.2 Tie-back connection

Tie-back connections have lower stiffness and yield strength; so, they govern the

force that can be carried by the cladding. Tie-back connections are designed to

provide a large level of ductility (Baird, Palermo et al. 2011). Ductile tie-back

connections allow relative movement between cladding and structure; also, they can

deform under in-plane loading while providing out-of-plane support (Metelli and Riva

2007). Tie-back connections must also be capable of withstanding out-of-plane wind

forces on the panel. They are remarkably weaker and less stiff than the bearing

connections and cannot carry sufficient force to ever yield; thus, the panel is assumed

to remain elastic. If a connection fails for any reason, it is preferable to replace the

faulty connection rather than the whole damaged panels (Khoraskani 2015).

3.3.3 Governing Failure Mechanism in Attachment

Cladding system is composed of structural frame members, a connector body, and a

cladding panel linked to each other with strong and stiff attachments. If the in-plane

strength of the cladding panel is greater than that of the connector body, then the

connector body (weakest link of the chain) is expected to govern the overall cladding

85

failure mechanism. On the opposite side, if the connector body is stronger than the

panel, then failure is governed by the panel strength and it is considered the weakest

link of the chain (Baird, Diaferia et al. 2011). For the two above-mentioned scenarios,

it is assumed that the attachment of the connector body is stronger than those of both

the cladding and the connector body itself (Toledo Arias 2013).

Figure 3-15: Different failure mechanisms and push-over behaviour of precast panels attached to a frame system (Baird, Diaferia et al. 2011)

It should be noted that, when the attachment governs the failure, then the risk of

falling panels is very high. This contribution is greater when panels are attached to the

columns, rather than beams, because the beams deflect more and activate the

connections later (Toledo Arias 2013). Different failure mechanisms, push-over

behaviour of precast panels attached to a frame system and seismic interaction

between precast concrete cladding systems and moment resisting frames are shown in

Figure 3-15 (Baird, Diaferia et al. 2011). Facade frames and window sashes tend to

86

move with local distortion during seismic activities, which result in a change in the

angularity of corners, known as racking, in buildings. Inter-storey drift, which is a

very crucial parameter for deflection and causing damage in structures, can result in

cracking of glass panels during earthquakes. AS 1170.4 (2007), clauses 5.4.4 and

5.5.4, specify that, "the inter-storey drift at the ultimate limit state, calculated from the

forces determined according to strength and stability provisions shall not exceed 1.5%

of the storey height for each level and the attachment of cladding and facade panels to

the seismic-force-resisting system shall have sufficient deformation and rotational

capacity. This requirement is for the ultimate limit state of buildings for seismic

performance and, for a typical 3600 mm heigh floor, it amounts to relative floor to

floor building deflection of 54 mm.

New Zealand Standard "Earthquake actions", NZS 1170.5, (2004) specifies in

clause 7.5 that, a maximum inter-storey drift limit of 2.5% is applicable for the

ultimate limit state for a 500 year Return Period event. In the case of a 2500 year

Return Period near fault event, this limit has to be increased to 3.75%. Drift limits of

2.5 and 3.75% create demands of 90 and 135 mm, respectively, on facade systems

while assuming the storey height of 3600 mm. Council on Tall Buildings, Group SB

(1979), examined the serviceability wind drift criteria from industry sources and

literature and found that drift limits ranging from 0.001 to 0.004 H (H: height of the

building) were used. However, the council stated that buildings designed in the past

have been known to perform satisfactorily when designed for drift limits from 0.002

to 0.005 H. ASCE Task Committee found that most of the designs for institutional,

commercial, and residential building types used drift ratios in the order of 0.002 to

0.0025 H for steel-framed buildings. Unitized curtain wall systems, with the increased

87

usage in glass, are particularly vulnerable during seismic activities and pose the threat

of flying and falling glass.

Numerous and extensive studies have been done after widespread glass damage in

Mexico (1985) and San Francisco (1989) due to devastating earthquakes (Hareer,

Environment et al. 2006). Two main factors should be evaluated when considering

earthquake damage:

How glass systems perform and respond to racking.

How glazing performs after being cracked from frame movement.

Despite extensive studies in the area of seismic loading and response, major

building codes do not address these issues. Areas that are considered in industrial

standards include:

American Architectural Manufacturing Association (AAMA 501.4) has

developed a test method, AAMA 501.4, focusing on serviceability, which does

not address life safety.

National Earthquake Hazards Reduction Program's (NEHRP) guidelines,

published by Federal Emergency Management Agency, cover life safety and

ensure that glass fall-out does not occur due to the movement of building frame.

International Building Code (IBC), 2003 edition, refers to seismic design

provisions contained in ASCE 7-02 (Minimum Design Loads for Buildings and

Other Structures).

ASCE 7-02, in turn, refers to AAMA 501.6-01 (Recommended Dynamic Test

Method for Determining the Seismic Drift Causing Glass Fallout from a Wall

System).

3.4 Façade Panels Capability and Compatibility to the Proposed Novel Designs

The proposed solutions for each facade system will be discussed case by case to

evaluate their potential for incorporation into newly designed connections. In Chapter

11 of FEMA356, Architectural components are categorized as glass blocks,

88

prefabricated panels, and glazed exterior wall systems (FEMA356 2000, FEMA-389

2004). Facade panels, including their possible damage and capacity to incorporation

into a new system, will be presented and discussed case by case.

3.4.1 Infills

Infill facades have usually the worst performance among the facade groups; indeed,

most of them are categorized as Operational or Immediate Occupancy after

earthquakes (AS1170.4 2007). Modern glazing infills generally perform much better

and do not suffer from as much damage as the older infills. Reinforced masonry infill

did not typically show much damage other than small cracks and it has been proven

before that the movable infill has a positive effect on the seismic performance of the

primary structure (Behr, Belarbi et al. 1995, B. Samali 2014, B. Samali 2014). The

infill facade panels that are parallel to earthquake action are like shear walls (shell

elements) inside each frame of the structure and can represent in-plane behaviour

during excitation (Abtahi, Samali et al. 2012).

During computer modelling, each panel was attached to the main structure frame

by two horizontal and vertical springs on each side. The horizontal springs have

specific stiffness and damping; but, the stiffness varies between some reasonable

values in order to evaluate the response of the main structure. The vertical springs are

stiffer, only carry the weight of the facade panels, and transfer gravity loads to the

structural elements. If a facade panel can withstand the internal applied force of an

earthquake and the facade glass does not break, then they can represent the infill walls

and increase the stiffness of the main structure (B. Samali 2014). This is why the

following three parameters must be considered in this proposal: using glass with

appropriate thickness; using high strength Laminated Safety Glass; and using a shock

89

absorber to dampen parts of the applied force; in fact, the thickness of glass for each

panel should be designed according to AS.4667 standard provisions. Laminated glass

is a composite material consisting of two or more sheets of glass that are permanently

bonded to each other by a plastic interlayer material. Since the relatively high surface

compression inherent in laminated glass is a crucial parameter when designing the

toughening process, it should be used to increase the strength of the glass and produce

fracture characteristics; however, this process can increase the risk of spontaneous

fracture.

Moreover, since tempered glass cools down faster while being heat strengthened,

compression on the surfaces and edges is higher. In fact, the surface compression is at

least 69,800 kN/m2 and edge compression is at least 67,700 kN/m2, which means that

tempered glass is 45 times stronger than annealed or untreated glass; therefore, it is a

reliable option for the proposed application. Another method of increasing the

consistency of glass panes is to use organic safety film or other glass coatings (Behr,

Belarbi et al. 1995)

3.4.2 Lightweight Cladding

Normal lightweight cladding weighs less than 80 kg/m2. Furthermore, 82% of

lightweight claddings is deemed to be either at Operational or Immediate Occupancy

levels and exhibit either no damage or very minor damage such as ejected window

seals or cracked glass after earthquake activities (Pinelli, Craig et al. 1995). Damage is

categorized by damage to the frame and glass, respectively. Newer lightweight

claddings are proportionately less likely to exhibit moderate to severe damage.

However, issues still exist with current design and construction techniques, because

90

even lightweight cladding systems of less than 20 years old experience heavy damage

(De Matteis 2005).

In some cases, screws which attach the facade frame to the sub-frame are either

sheared off or torn out of the wood (Palermo, Pampanin et al. 2010) . During an

earthquake, multiple sections of the curtain wall become completely detached from

the building and, in extreme cases, the aluminium frames and glazing fall onto the

footpath (Georgiou 2010). The main reason why glass breaks is because insufficient

allowance is made for the movement of glass panels during an earthquake. Glass

damage is common in most cases where lightweight cladding contains glass; so,

damage is categorized according to FEMA 356 (FEMA356 2000). These systems are

sub-categorized as shown below and their ability to be incorporated into the proposed

system will be discussed case by case

3.4.3 Curtain Walls

Curtain wall facades are designed to span multiple floors; so, they only carry their

own self weight; however, they are designed to resist sway induced by wind

and seismic forces acting on the building (AAMA 501.4 2000). A horizontal wind

load is transferred through the wall and then to the main building structure through the

anchors which attach the mullions to the floors or columns of the building (Behr

1998). In most situations, the curtain wall can naturally withstand seismic and wind

induced building sways because of the space provided between the glazing infill and

the mullion. Curtain wall systems are designed to accommodate 25-35 mm of relative

floor movement (vertical movement) without overall system performance failure. But

sometimes large floor drifts transfer large forces to the façade brackets and may result

in panel damage, glass breakage, or falling of the whole frame. Stick systems that act

91

like curtain walls are a desired lightweight choice in modern multi-storey buildings

(Chew, Tan et al. 2004). They are installed as long pieces (referred to as sticks)

between floors vertically, and between vertical members horizontally. Unitized

Curtain Walls, as newer systems, are hung on the structure of a building as an

enclosure (Behr, Belarbi et al. 1995). They have a similar configuration to a stick

system but they are much faster to install because most of their fabrication is carried

out in a factory (Memari, Behr et al. 2003). These systems cannot be incorporated

with an in-plane concept to use their in-plane stiffness as additional stiffness for the

whole structure because they are not located inside the main frame of the structure,

and they also have a low allowable movement (only 75mm) so incorporating them

with out-of-plane concept using anchorages with very low axial stiffness is not

practical either (Sanders 2006). Indeed, as proven before, façade frames with low axial

stiffness in brackets have excessive back and forth movements, which means

widespread damage and casualties may happen (Pinelli, Craig et al. 1995)

3.4.4 Panelised Curtain Wall

It was shown previously that numerous severe damage is likely in the spider

glazing (Bolted Glass Façade) after an earthquake excitation (Amadio and Bedon

2012), with the damage initiating around the “spider” that holds each panel of glass,

possibly because the spider concentrates the stress in these regions due to the

connection being attached to the structure. The possibility of incorporating bolted

glass facades into the proposed system to reduce the response of primary structures to

earthquakes is very low (Roth 2007, Permasteelisa 2009). Indeed, it has been proven

that, if some design changes are made and viscoelastic spider connectors are used in

the cable supporting facade system, they could mitigate the force of an air blast load

92

applied to the main components of facades, especially glass panels; but, during

earthquake excitation, which has a longer applied force and different range of

frequencies, the scenario is completely different (Walker III, Niemoeller et al. 2011).

3.4.5 Double Skin Façade System

One of the more recent variations to the stick system is a double-skin facade system

that consists of two layers of facade material (typically glass) which creates a sealed

cavity and improves the thermal and acoustic performance of the building (Pinelli,

Craig et al. 1995). Double-skin facade systems are being increasingly used in high-

profiled buildings and, because of their configuration and the gap between the two

layers, they have the capability of being incorporated into the newly proposed system.

Using the in-plane stiffness concept in the inner layer of double-skin facade is a good

option; but, it needs some changes in design and the glass panels need to be stiffer.

Moreover, the outer skin can be attached to the inner frame and/or primary structure

with low shear stiffness brackets that can move freely during an earthquake, and the

gap between each panel must be large enough to avoid any extra damage when the

panels collide. The only barrier for the application of the in-plane concept in double-

skin system is small mass of the outer skin in comparison to overall mass of the main

structure (Arons 2000).

3.4.6 Heavyweight Cladding

Heavy claddings perform better than most facade systems and perform at either

Operational or Immediate Occupancy ranges because they have exhibited little to no

evidence of damage. The performance level of Immediate Occupancy is used for

claddings that show some evidence of cracking or where it was clear that the panels

had residual displacements and/or rotations. The performance level of Life Safety is

93

used for claddings where damage is visible; in some cases, several spandrel panels

shear off their bolted connections and fall onto the footpath below; this type of

damage is classified as high hazard (Baird, Diaferia et al. 2011).

A complete disconnection of panels, which leads to racking and spalling, is also

common, as is minor damage in the form of panels with residual displacements and/or

rotations and ejection of sealing joints between them. After considering the weight of

these panels, using the newly proposed system would be risky and requires great

attention to details and application of higher safety factors (Hunt 2010). It is shown

that facade panels could be considered as structural elements and movable panels.

Also, they have the potential to reduce building lateral displacement, if axial stiffness

of facade brackets is tuned to the first frequency of the main structure during

excitations (Sacks, Eastman et al. 2005).

But, it is proven that, due to low calculated axial stiffness of panels' brackets, they

have excessive lateral movements, which demonstrates a significant barrier for the

practical implementation of the proposed system. In addition, frequency content of

earthquake records is not identical and may detune the calculated stiffness of brackets

(Behr and Belarbi 1996). Several research groups have investigated the interaction

between precast cladding panels and the supporting framing. Analytical and

experimental studies have revealed that cladding may have a significant influence on

the seismic response of the building as a whole (Yee and Eng 2001). Little to no

damage is expected in the precast panels themselves due to their thickness and

rigidity. As such, the panels are modelled to behave as rigid blocks, and the damage to

the cladding system is concentrated on the connectors and window glazing system

(Goodno and Palsson 1986).

94

3.5 Chapter Summery and Conclusion

The effects of earthquake load on mid-rise and high-rise buildings are well known by

the engineering community. In the case of this kind of extreme loading, building

structures may be susceptible to excessive deflections and noticeable accelerations and

such problems are usually reduced by the adoption of external damper systems. The

loss of valuable and prime space coupled with the initial cost of installing large sized

damper systems have been accepted by building owners with reluctance and any

viable alternative system to dissipate seismic excitations would be highly appreciated

by them. Similarly, for medium sized buildings the reliance on damper systems to

alleviate earthquake induced actions has also received much attention in recent years.

To date, very few engineers and architects have exploited the potential of façade

systems as an energy absorbing system to combat seismic loads. Most attempts, so

far, have considered façade as an add-on with no or little structural contribution and

this is evidenced by the exclusion of façade systems in computer modelling of

building structures as an analysis tool. Double-skin façade systems (DFS) are

becoming very popular for improving the sustainability of commercial buildings in

Australia and overseas. In these cases, it is proposed to use a moveable façade skin

attached to passive devices which are, in turn, attached to the main building frame.

The energy imparted by earthquake forces can be dissipated by the absorbing façade

system with major economic benefits. It is important to mention that Double Skin-

Façade (DSF) system is the selected type of façade system to be used in this thesis.

This kind of façade system has an adjustable cavity in which the damper device can be

installed and this gap allows for in-plane and out-of-plane movement of outer skin of

DSF system during earthquake actions.

95

CHAPTER 4

FEASIBILITY STUDY &

PRIMARY NUMBERICAL

MODELLING

96

4.1 Introduction and Methodology

To date, the engineering community has viewed structural facade systems as non-

structural elements with a high aesthetic value and a barrier between the outdoor and

indoor environments. The role of facades in the dissipation of imparted energy due to

external loading in a building has been also recognized and the industry is also

witnessing the emergence of many energy-efficient facade systems (Abtahi, Samali et

al. 2012). It has been also recognized that facade systems add some stiffness and

damping to the overall building despite the new and modern systems; e.g. curtain

walls which add a relatively small amount of stiffness and damping to the overall

building (Mohotti, Lunmantara et al. 2013).

Despite these advancements, facade has rarely been considered or designed as a

potential earthquake-induced vibration absorber for structural buildings (Houston

2011). The potential of utilizing a movable exterior skin in a typical double-skin

facade system is investigated in this chapter and shown that, with optimal choice of

materials for stiffness and damping of brackets connecting the two skins, a substantial

portion of earthquake-induced vibration energy can be dissipated in order to avoid

expensive aseismic designs. The initial work has demonstrated that the seismic

response for mid-rise buildings subjected to moderate earthquakes can be substantially

reduced by introducing a smart design of a double-skin system connections.

During a significant seismic activity, glass openings often break and also

aluminium transoms and mullions are distorted badly. From architectural and

economical viewpoints, damage to these expensive non-structural elements causes

great economic losses to building owners. Architects, engineers, and designers have

made great efforts to minimize frame and glazing concerns based on the established

97

performance criteria and standards during recent decades (Yankelevsky, Schwarz et

al. 2011). But, with stiff connectors, which attach facade panels to floor slabs, this

problem still exists. Glazing panels should be designed in a way to move backward

and forward when subjected to external forces; therefore, they perform better under

these circumstances. This suggestion not only decreases potential losses and mitigates

damage, but also reduces glass brakeage and injuries to occupants from fragments

(Sivagnanasundram 2011).

The primary idea for the design of advanced connections is to reassign a structural

role to the architectural facade in order to introduce the added passive damping into

the structural system. Damage assessment of facade frames and threat of glazing

thrown to the ground will be evaluated as the possible stages of analysis in order to

ensure that acceptable protection is considered and design remains within the available

funds. The outcome of this study would lead to the development of a new generation

of facade bracket elements, which can sacrifice themselves to absorb some parts of the

applied energy of earthquake and reduce seismic response of the main structure for

new as well as the existing buildings to be retrofitted (Li, Hutchinson et al. 2011).

As described earlier, analytical analysis in this research is divided into three

different sections as described below. Generally, there are two design criteria in facade

design in order to attain the desired level of performance; the first is their ability to

withstand environmental forces such as earthquake and severe winds and, second, they

can be designed to break away in order to avoid overloading the structure (Elghazouli

2009).

4.2 Behaviour of Double-Skin-Façade in Suppressing Earthquake Loads

4.2.1 Introduction

In recent decades, buildings with significant usage of glass are becoming more

common. The development of non-load bearing curtain walling technology around the

turn of the 20th century, along with double skin façade (DSF) system, which have

substantial cavity space between the inner and outer façade layers, have increased

interest in these systems with the aim of fully exploiting their potential. Building

façades generally perform as environmental medium between the controlled interior

and harsh exterior as well as building identifiers through their aesthetic design.

Kareem (1992)) proposed the concept of isolation in the mountings of the cladding

to the structural system. Buildings are isolated from earthquake excitation by

employing isolator bearings between the building and the foundation and a similar

concept is proposed for cladding. If the cladding is connected to the frame by an

isolation mounting, then the dynamic loads transferring from the frame will be

reduced and consequently the building motion will be reduced as well. In order for

this mounting to be effective, the ratio of excitation frequency to the natural frequency

of the cladding should be greater than square root of two (Kareem 1992). In this

situation, the mounting system is more effective without any damping.

The proposed system can be materialized by dividing the cladding on the building

envelope into several segments. The preliminary calculations of Kareem (1992)

suggest that such a mounting system will be quite soft and pneumatic mounts may be

an appropriate choice here. Such an installation may cause the cost of a cladding

system to be, however, very high. This can be overcome by using these systems in

staggered configurations and the remaining portions of the building envelope may

98

utilize conventional cladding. Moon (2005)) showed that dynamic motion of tall

buildings can be reduced, for example, by more than 50% when the DSF façade

connectors are designed to have about half of the primary structure frequency.

However, there exists a design challenge which is the excessive and extreme motion

of the DSF outer skins, which would disturb occupants through visible cues, and

would potentially undermine the ventilation system intended by DSF systems through

pumping cavity air around the building.

4.2.2 System Modelling

To evaluate the seismic behaviour of the proposed system, the main structure and

the facade system were simplified and modelled as a two-degree-of-freedom system as

shown in Figure 4-1. The system consisted of a primary mass (m) of the main building

structure along with the secondary mass (md) corresponding to the facade system. The

concept of the simplified system model is in essence similar to the concept of a tuned

mass damper (TMD) modelled in structures. The two masses were attached by a low-

axial-stiffness spring ( ) and a damper system (Moon 2009).

Figure 4-1: Simplified model of the primary structure and façade system connected by movable brackets

Connection properties concerning stiffness and damping were evaluated and varied

to achieve the appropriate response. In order to achieve optimal performance of the

99

100

proposed system, the connection frequency was tuned to the primary mass frequency

of the structure. Dynamic force was applied to the main mass and, through the

connections between the primary and secondary mass, was transferred to facade

frames. The outer skin mass is assumed to be around 2% of the primary structure

mass. Details of facade panel attachment to the main structure and how it is modelled

as a simply supported beam in computer modelling are presented in Figure 4-2.

Figure 4-2: Detail of façade connection to primary structure and modelling assumption in SAP2000

4.2.3 Dynamic Responses of the System

Below are the governing equations of the system shown in Figure 4-1:

(4-1)

(4-2)

where primary structure mass; m DSF outer skin mass; k primary

structure stiffness; k DSF connector stiffness; c primary structure viscous

damping parameter; c DSF connector viscous damping parameter; p applied

dynamic loading; u primary structure maximum lateral displacement; and u DSF

outer skin maximum lateral displacement. It is convenient to work with the solution

expressed in terms of complex quantities. The force is expressed as:

m

101

(4-3)

Where forcing frequency and is a real quantity representing the loading

amplitude. The response is taken as

(4-4)

(4-5)

Where natural frequency of the primary structure, and

the response amplitudes, and , are considered to be complex quantities. Then the

corresponding solution is given by either the real or imaginary parts of and .

Substituting Equations (4.3)–(4.5) into the set of governing Equations (4-1) and (4-2)

results in:

Ω Ω Ω (4-6)

Ω Ω (4-7)

Considering the following notations:

(4-8)

2 (4-9)

where primary structural damping ratio, and

(4-10)

p

u fu

u fu

102

and natural frequency of the DSF outer skin, Stiffness of the brackets which

is a variable and a function of the input frequency.

2 (4-11)

where façade connector damping ratio. Defining as the DSF outer skin mass

to primary mass ratio

(4-12)

and defining f as the DSF outer skin frequency to primary structure frequency ratio,

(4-13)

and defining as the forcing frequency to primary structure frequency ratio,

ρ (4-14)

Then the corresponding Frequency Response Functions (FRF) can be obtained by

derivation from the equations of motion as follows.

(4-15)

(4-16)

4.3 First Proposal (Feasibility Study of Façade System as Multi Tuned Mass

Damper with 3D Numerical Modelling in SAP2000)

As described before, effectiveness of a TMD for the response reduction of a single

degree of freedom (SDOF) system could be simply extended to continuous MDOF

structures such as tall buildings by a modal approach. A single TMD cannot provide

vibration control to more than one mode; therefore, Multi-Tuned Mass Dampers

(MTMDs) have used to evaluate and control multiple structure modes of vibration (Li

f fk

f m

2)322223232(2)242222422(

22244

ffmfffffffffmf

fffH

2)322223232(2)242222422(

2242)12(

ffmfffffffffmfmfH

103

2002). Initially, a comprehensive literature review was required to evaluate in depth

the feasibility of using movable facade systems as a Multi Tuned Mass Damper along

the height of the structure. Therefore, numerous studies and various computer model

evaluations were conducted based on the design procedure of a TMD system in the

analytical part of this research to evaluate the simulations. The ratio of facade mass to

main structure mass was a main issue in this evaluation for using TMD design

procedure to achieve correctly tuned damper properties and facade mass relative to the

main structure. At the first stage, to show the feasibility of the movable facade

concept, SAP2000 structural analysis package was used to provide advanced nonlinear

time-history analysis on a 3D frame structure with and without moving facade systems

subjected to two near-field and two far-field benchmark earthquakes used for the

assessment. This program was used to generate the geometry, boundary conditions,

and loading conditions of the model as well as analysis.

4.3.1 Earthquake Loading Records and Boundary Condition:

Earthquake loads are normally defined as lateral dynamic loads and can be vertical

loads as well. These loads are very complex, and potentially more destructive than

wind loads. In an earthquake-prone zone, every structure must be designed to

withstand and survive damaging earthquakes. The scaling of earthquake loads are

based on time domain and frequency content of the records are unchanged during the

whole record. The seismic loadings applied to the structural models in this chapter

were from time histories of horizontal ground accelerations of chosen past earthquake

records. The accelerations were applied in the x-direction at the base of the structure.

Four typical earthquake records suggested by International Association for Structural

104

Control and Monitoring including two far-field and two near-field earthquake

excitations were chosen for the primary study.

Table 4-1: Earthquake ground motions used in this study

Earthquake Country Year PGA(g) Mw(R)Northridge USA 1994 0.843 6.7El-Centro USA 1940 0.349 6.9

Kobe Japan 1995 0.833 6.8Hachinohe Japan 1968 0.229 7.5

These records were El-Centro (far field), Hachinohe (far field), Northridge near

field), and Kobe (near field) earthquakes which are shown in Figure 4-3 to Figure 4-6

and Table 4-1.

Figure 4-3: Scaled Northridge earthquake excitation record

Figure 4-4: Scaled El-Centro earthquake excitation record

Figure 4-5: Scaled Kobe earthquake excitation record

-1-0.8-0.6-0.4-0.2

00.20.40.60.8

1

0 5 10 15 20 25 30

Acc

eler

atio

n (g

)

Time (Sec)

-1-0.8-0.6-0.4-0.2

00.20.40.60.8

1

0 5 10 15 20 25 30 35 40 45 50 55

Acc

eler

atio

n (g

)

Time (Sec)

-1-0.8-0.6-0.4-0.2

00.20.40.60.8

1

0 5 10 15 20 25 30 35 40 45 50

Acc

eler

atio

n (g

)

Time (Sec)

105

Figure 4-6: Scaled Hachinohe earthquake excitation record

To confirm the effectiveness of the façade systems in the range of frequencies of

interest, it is critical to establish the frequency range where the maximum seismic

energy is concentrated. Therefore, elastic response spectra of the four earthquake

records are computed by means of the Newmark time integration method (Streicher,

Heimrath et al.) by using PRISM software (Streicher, Heimrath et al.). By using this

program seismic response analysis of SDOF system can be calculated. Response

spectra of the earthquake records are shown in Figure 4-7, Figure 4-8, Figure 4-9 and

Figure 4-10.

Figure 4-7: Response Spectra of scaled Northridge earthquake record

-1-0.8-0.6-0.4-0.2

00.20.40.60.8

1

0 5 10 15 20 25 30 35

Acc

eler

atio

n (g

)

Time (Sec)

0

0.5

1

1.5

2

2.5

3

3.5

4

0 5 10 15 20 25 30 35 40 45 50

Acc

eler

atio

n [

g]

Frequency (Hz)

106

Figure 4-8: Response Spectra of scaled El Centro earthquake record

Figure 4-9: Response Spectra of scaled Kobe earthquake record

Figure 4-10: Response Spectra of scaled Hachinohe earthquake record

From these spectra it is clear that most of the seismic energy is concentrated

between about 2 and 5 Hz and, therefore for the façade system to act as an effective

damper system it should possess a system frequency in the same range. In order to

consistently compare the response of a structural model under different earthquakes,

0

0.5

1

1.5

2

2.5

3

3.5

4

0 5 10 15 20 25 30 35 40 45 50

Acc

eler

atio

n [

g]

Frequency (Hz)

0

0.5

1

1.5

2

2.5

3

3.5

4

0 5 10 15 20 25 30 35 40 45 50

Acc

eler

atio

n [

g]

Frequency (Hz)

0

0.5

1

1.5

2

2.5

3

3.5

4

0 5 10 15 20 25 30 35 40 45 50

Acc

eler

atio

n [

g]

Frequency (Hz)

Hachinohe

107

these earthquake records were scaled to have identical peak ground acceleration

(PGA). The supports at the base of the structure were modelled as rigid, restrained

against translation and rotation in x, y, and z directions. The vertical loading on the

structure was in the form of uniformly distributed loads applied to the beams.

4.3.2 Material Properties

Structural materials used in the computer modelling are listed in Table 4-2 and

Table 4-3. These values are obtained from the properties of common materials used by

Australian construction companies.

Table 4-2: Material properties of façade system

Façade Sections Modulus of Elasticity

(kN/mm^2) Poisson’s Ratio Density (kN/m^3)

Glass 62.76 0.2 22Aluminium 68.64 0.33 27

In Table 4-2, module of elasticity, Poisson's ratio, and density of materials used for

facade components in SAP2000 modelling are shown. Aluminium was used to

construct the frame connecting directly to the glass facades via rubber sealants.

Rubber used as a sealant at the junctions between the aluminium frame and glass

facades is not modelled in the preliminary structural analysis. Typical values for

compressive strength of concrete, used as different components of the structure are

based on strong column and weak beam in the aseismic design of reinforced concrete

structures. These values are shown in Table 4-3.

Table 4-3: Material properties of primary structure

Compressive Strength (MPa)

Modulus of Elasticity (GPa)

Poisson’s Ratio

Density (kN/m^3)

Column 6030 0.2 24.5Beam 40

Slab 35

108

4.3.3 Structural Modelling and its Dynamic Behaviour

To show the feasibility of the proposed concept, a 3D 10-storey concrete moment

resisting building frame with the total height of 42 m and width of 18 m consisting of

three spans in each direction is selected. Structural plan of the model is shown in

Figure 4-11 .

Figure 4-11: Schematic plan of the structural model

Span length of each bay is 6.0 meters in each direction; the height of the first floor

is considered to be 6.0 m to represent podium facade and the other stories are 4.0 m

in height. As illustrated in Figure 4-12, the distance between the facade panel and

primary structure, called cavity, is considered to be 200 mm to provide enough

space for the installation of dampers. As can be observed, the facade panels are

attached to the primary building by some brackets. These brackets are assumed to

be flexible in order to absorb maximum external energy induced by the earthquake

and are considered as a damper element in SAP2000 modelling. The facade system

is attached to the ground by a pinned joint at ground level in order to allow them to

rotate freely during seismic analysis. The maximum number of damper elements

considered on all sides and levels of the main building is shown in Figure 4-12. The

system is fully attached to the main structure at levels zero, one, five, and ten to

maintain the stability and proper connection of facade panels to the primary

109

structure. Attachment between facade panels on each floor is assumed as a pinned

joint and they are attached to each other from the above and below by these pinned

joints to move freely during the application of environmental forces. The gap is

called stack joint and is located 900 mm above the floor slab level on each floor.

The stack joint is designed to resist lateral loads, while the two floor anchors resist

both lateral loads and applied gravity load to the concrete slab. Time-history

dynamic analysis is selected to acquire the response of the structure under seismic

loading. primary structure’s damping ratio is assumed to be 5%, which is within the

range of statistically reasonable values based on the measured natural damping

ratios for mid-rise building.

Table 4-4: Structure model dynamic properties

Weight (Knaack, Klein

et al.) Period (Sec) Frequency(HZ)

Bare Frame 2,800 0.78 1.28

Frame with Movable facade

2,870 0.97 1.03

This analysis assembles the mass, stiffness, and damping matrices and solves the

equations of dynamic equilibrium at different time instances. The response of the

structure is obtained for the selected time steps of the input earthquake record. The As

shown in Table 4-4, the vibration period of the bare frame is slightly less than the

building with movable facade system. Dynamic properties of damper system are very

crucial for evaluating the potential of the movable facade system. Damping ratio is

considered to be 5% of critical for each of the damper elements. Stiffness of the

damper elements is set in a way to act as a multi-tuned mass damper along the height

of the structure and their main frequency is tuned to the first or second frequency

mode of the primary structure.

110

Figure 4-12: Schematic view of primary structural model with details of brackets and connections

200 mm

111

4.3.4 Results of Computer Modelling

Effects of implementing the maximum number of damper systems, which can be

installed in a 10-storey structure system, are evaluated. These dampers are installed

between the main structure and glazing system on two faces of the building, along the

same direction as the applied earthquake excitation. Totally, 28 cases are investigated

with the stiffness of the connections in the range (5-10,000 kN/m). The optimal value

of stiffness was found to be around 10 kN/m for all four earthquake records. The total

weight of the structure including cladding panels is 2,870,kN. The results of the above

analyses are then compared with those of the structural system without energy

absorbing connections. Time-history responses of top floor displacement for the

structure without facade frames, structure with ordinary rigid facade connections, and

structure fitted with advanced connectors under the selected earthquake excitations

obtained from analytical models using computer program SAP2000 are illustrated in

Figure 4-13 to Figure 4-16.

Figure 4-13: Time history analysis of structural model under Northridge earthquake

Figure 4-14: Time history analysis of structural model under El-Centro earthquake

-800-600-400-200

0200400600800

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Bare frame Conventional façade Movable façade

-800-600-400-200

0200400600800

0 3 6 9 12 15

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m)

Time (Sec)


Figure 4-15: Time history analysis of structural model under Kobe earthquake

Figure 4-16: Time history analysis of structural model under Hachinohe earthquake

It can be seen from the figures and Table 4-5 that, with suitable damper connection

properties, acceptable reductions in lateral displacement of the building under all

earthquake excitations can be achieved. It is observed from these figures that the

facade system, even a fixed one, can slightly reduce the top displacement of structural

models in certain situations.

Table 4-5: Results of time history analysis

Records Peak Displacement (mm)

% Reduction in Peak Displacement Without

Facade With Fixed

Facade With Movable

Facade Northridge 492 525 356 27 El-Centro 333 298 235 25

Kobe 436 282 196 54 Hachinohe 273 260 206 25

However, the fixed facade system slightly increases the response of the main

structure during most parts of the Northridge earthquake. According to these results, it

is seen that, with the use of movable façade, the overall lateral displacement of the

primary structure, subjected to seismic loads, is somewhat decreased. The seismic

response of the building facade system, under Northridge earthquake, is higher in

-800-600-400-200

0200400600800

0 3 6 9 12 15

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t(m

m)

Time (Sec)


-800-600-400-200

0200400600800

0 3 6 9 12 15

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t(m

m)

Time (Sec)


112

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value than the El-Centro results. However, better results with higher response

reduction are also achieved under Northridge earthquake, considering this system in

facade connections. As demonstrated in El-Centro, Hachinohe, and Northridge, the

reduction is almost similar and there is around 26% reduction in lateral displacement.

In Kobe earthquake, this percentage is much higher: i.e. 54 %. As illustrated in Figure

4-16, after the second peak, which is the biggest in the time-history analysis,

effectiveness of the damper is reduced too much or even is amplified in the negative

direction. This phenomenon happened in Kobe excitation because of jerky behaviour

of this earthquake that could lead to detuning of damper system as a result of the

effect of very high dynamic impact load to the primary structure after the second peak.

Comparison of responses for the undamped and damped structures demonstrates that

the advanced bracket connectors are able to decrease the peak values of top floor

displacement of the primary structure.

4.3.5 Conclusion:

After numerous time-history analyses in SAP2000, the concept of the integrated

façade/damper system is proven promising. However, there is some ambiguity about

some parameters and assumptions in the first part of this study, as follows:

By using bracket elements with low axial stiffness, the outer skin starts to move

excessively during earthquake excitation. Excessive movement of panels beyond

the allowable practical limits is not acceptable.

Facade panel mass should be considered based on the reality and practical values

which is generally small in modern buildings.

The damper device with high damping capacity should be selected.

The external skin of double-skin facade as a TMD system does not seem to be

effective, because the overall mass of the facade panels is not enough to be

considered as MTMD system along the height of the structure in order to affect the

primary structure response.

4.4 Second Proposal - Numerical Modelling of Facades with Sacrificial

Elements in SAP2000

4.4.1 Introduction

With well-designed structural connections, plastic hinges will form in beam elements

during a severe earthquake; they absorb earthquake energy and change frequency of the

structure as well. Based on the aforementioned design, the preferred order of failure in a

well-designed structure should be in this order: beams<connection<columns. At the

second stage, a 2-D structural model is evaluated and natural damping of bare frame is

compared to the structure which is equipped with damper elements between facade and

the primary structure. Therefore, there are two primary aims here:

The possibility of bringing the response of main structure to an acceptable limit

with the maximum number of dampers

Study the effect of maximum number of dampers on the overall damping behaviour

of mid-rise structures

In following configurations, out-of-plane and in-plane behaviours of facade panels

are evaluated to show the possible effect of bracket configuration on the primary

structure's lateral displacement. The analysis is conducted to determine the sensitivity

of parameters such as facade panel height, damper locations, number of dampers, and

damping of supporting bracket systems in achieving an optimal system. The selected

loading is again the typical earthquake records suggested by International Association

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115

for Structural Control and Monitoring. All the numerical simulations were conducted

in SAP2000 program. For the initial studies, El Centro earthquake is applied to

evaluate the response of the whole system in different situations.

4.4.2 Preliminary Numerical Modelling

According to the results of the previous sections, it is concluded that design

procedure for bracket elements should be changed to have meaningful reductions of

primary structure displacements. To achieve this, the idea is to define a bracket

element which has elasto-plastic behaviour. Their initial stiffness is similar to the

practical values which is around 40 . The first innovation is to define plastic hinges

in the brackets with different values of plastic force plateau to evaluate their

effectiveness during dynamic excitations. By creation of these plastic joints the

brackets can vibrate like multiple pendulums and better dissipate the seismic energy

without total collapse. The large available ductility of the joints can absorb a large

portion of seismic energy and reduce lateral displacements. For this purpose, a 3-D

12-storey concrete structure consisting of four bays with the span length of 8 m in X

direction and five bays with the span length of 6 m in Y direction is selected. The

storey height is set at 3.0 meter.

Table 4-6: Frequency of first three modes of the structure

Modes Natural Frequency(Hz) Period of Vibration(T in Sec) 1 0.92 1.092 3.45 0.343 5.26 0.19

Based on Australian and New Zealand standard (AS/NZS 1170.1: 2002) permanent

and imposed loads were calculated and applied to the structures in the form of

uniformly distributed loads. The vertical loading on the structure was 40 kN/m applied

116

to the0020storey beams. Three different kinds of structures were evaluated. The first

one which is shown in Figure 4-17 is a structure with an elastic behaviour in which

dimensions of structural elements are designed and calculated so that only minor

damage is sustained during earthquake excitation. Dimensions of columns and beams

are 60x60cm and 60x65cm, respectively. Frequencies of the first three modes of the

structure are shown in Table 4-6.

Figure 4-17: Elastic structure in X direction

The second is a structure with pre-defined plastic hinges in beams and columns and

facade system with fixed bracket elements representing conventional facade system as

shown in Figure 4-18. Design of the building structure ensures that plastic hinges are

formed in beam and column structural elements (beam elements have higher value of

ductility compared with column elements). Based on these pre-defined plastic hinges,

the structure behaves plastically during earthquake excitation.

117

Figure 4-18: plastic structure with auto-defined plastic hinges in elements

Figure 4-19: plastic structure incorporated with façade elements with auto-defined plastic hinges in structural elements and user-defined plastic hinges in bracket elements

118

By introducing dissipating bracket elements, it is aimed to reduce formation of

plastic hinges in structural elements (reduce damage). Part of the applied load is

transferred to bracket elements and results in deformation of it and finally dissipating

the energy. The third structure shown in Figure 4-19 is similar to the second one plus

additional plastic hinges that are defined with different plastic plateau forces in each

facade bracket element. The main aim of this part is to evaluate the extent of top

displacement reduction of the primary structure with these elements by facilitating to

transfer some plastic hinges from main structural elements (beams and columns) to

these sacrificial elements. It is assumed that the brackets are replaced with the

sacrificial elements in computer modelling. Stiffness and weight of the elements are

exactly similar to those of real brackets in facade installation. This behaviour is

consistent with out-of-plane behaviour of facade panel brackets and the defined plastic

hinges are formed only based on axial forces. The occurrence of plastic hinge in the

brackets is based on the applied axial force from ground excitation which is applied to

the structure and then transferred to the brackets.

4.4.3 Out-of-Plane Concept of Façade Behaviour

In out-of-plane façade behaviour, the panels perpendicular to applied earthquake

direction are affected but the façade movement is parallel to the earthquake force.

Plastic hinges are formed based on the applied axial force to the bracket; i.e. if the

applied force is more than a pre-assigned value of force in force–displacement

relationship, then the hinges would start to form and absorb energy. Dimensions of

facade panels are typically 1.5m by 4m. One side of a building showing 8 panels

subjected to wind loads is depicted in Figure 4-20. According to the Australian Wind

code, AS1170.2, calculated ultimate design wind pressure is 1.5 kPa.

119

This calculation is based on height of the structure, topography and geography in

which the assumed structure is located, shielding and many other items which are

described in the code comprehensively. This means that around 9kN is applied to each

façade panel. Allowing for a factor of safety of 1.5, this value increases to 15 kN per

panel. It was assumed that each façade column (vertical elements modelled in the 2-D

numerical computer model) represents four façade panels in reality in terms of mass

and dynamic characteristics.

Figure 4-20: Typical façade panel subjected to wind forces

Based on below calculations, 60kN, as a rational value, is selected as the plastic

plateau force for the brackets. Weight of each panel is estimated as about 300 kg for

the simulation.

1.5Kpa x 1.5m*4m=9 kN

Safety Factor=1.5

Minimum axial load per panel= 9*1.5=15 kN

In order to simplify the numerical modelling and avoid time-consuming computer

modelling it is assumed that the dissipative bracket systems are located only in

120

location of structural column on the main building. Therefore, each bracket on each

floor is assumed to carry the weight of 4 facade panels in numerical modelling.

Minimum axial load per bracket: 15 kN*4 =60 kN

4.4.3.1 Results and Discussion

Results of the top lateral displacement of the main structure with different plastic

plateau forces (PPF) in facade bracket elements are shown in Figure 4-21. Comparison

of the top peak displacement is one of the most common ways for showing the

effectiveness of the proposed façade system. However, it has been seen that, in some

cases, just comparing the peaks is not reliable enough to express a measure of the

global improvement on the time-history.

Figure 4-21: Top lateral displacement of structure with plastic brackets in El-Centro earthquake

The root mean square, also known as square root of the quadratic mean, is

a statistical measure of the magnitude of a varying quantity. For zero mean signals, it

matches the standard deviation of the value distribution. It is especially useful

when variations are positive and negative like earthquake excitation. This parameter

gives better indications of the global behaviour of time-histories and it indicates how

the values are spread around the mean value. Root Mean Square (RMS) values of top

floor displacement subjected to El-Centro earthquake for all the modelled cases are

shown in Table 4-7.

-200

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200

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Conventional façade PPF=60kN PPF=30kN PPF=3kN

121

Table 4-7: Root mean square of top floor displacement cases with different plastic plateau forces (mm)

Conventional facade system

PPF 60kN PPF 6kN PPF 3kN PPF 0.5kN

106.61 106.37 106.16 102.35 109.10

4.4.3.2 Conclusion

Facade panels have been considered as non-structural elements in all simulations,

mainly because of their mass ratio in comparison to main structure and their role in

carrying gravity loads. After the preliminary analysis, it is proven that this mass is not

large enough to activate the bracket elements. Also, due to the earthquake excitation,

the structure's maximum axial force in each bracket element is about 6 kN which is far

less than 60 kN, calculated based on practical considerations. A very small value of

plastic plateau force (around 3 kN) should be considered in order to achieve a minor

visible, but not significant, effect on the top lateral displacement of the main structure.

It can be concluded that façade elements undergoing out of plane movements are not

effective.

4.4.4 In-Plane Concept of Façade Behaviour

Ideas of glass structures or load bearing glass walls have been proposed before

(Wellershoff and Sedlacek 2003). An example of this ‘engineered transparency’ in

building envelopes such as the Reinbach Pavilion is shown in Figure 4-22.

Figure 4-22: Rheinbach glass museum, Rheinbach (Wellershoff and Sedlacek 2003)

Facade panels moving in parallel to direction of earthquake motion can be considered

as shell elements inside each frame of structure to represent their in-plane behaviour

during seismic excitation. (Pinelli et al. 1992) undertook parametric studies on a 6-

storey steel frame building fitted with two precast cladding panels per bay. They

studied the incorporation of metallic dampers into the connectors used to attach

architectural cladding to a building. The studied structure was a ¼ scale 6-storey 3-bay

moment resisting steel frame building constructed in the 1980s for laboratory testing

at National Centre for Earthquake Engineering Research in the US. For the cladding

to-frame interaction, the test frame was provided with two precast cladding panels per

bay (Pinelli, Craig et al. 1995). The panels were considered to be rigid and each panel

was attached at its bottom to the steel frame by two rigid bearing connectors and at its

top by two advanced connectors, which was a metallic hysteretic damper. In this

section, similar to the previous research, the in-plane behaviour and movement of

double-skin facade panel is assumed to be integrated into the structural frame and is

modelled as a wall system in which the in-plane stiffness is incorporated into the main

structural frame. Stiffness of each vertical connection is calibrated in order to

represent the behaviour of the panels. The modelled spring element has only axial

122

stiffness with zero shear stiffness. Instead of modelling the glass and aluminium frame

of the facade panels, it is assumed that the shell element is made of a smeared material

with combined properties of both materials.

Figure 4-23: In-plane concept of façade behaviour (façade as a shell element in structure frame and its connections)

Stiffness of the shell element is chosen to be a practical value between glass and

aluminium in computer modelling. The length of structural bay is 8000 mm and it is

assumed that dimensions of each shell element is 7600mm by 3600mm by 200mm as

there is a 200mm gap distance between the structural frame and the shells in each

direction. They are attached to the structural frame on each of the four sides by

vertical bearing connections which are modelled as simple linear spring elements with

a very high stiffness value in the computer program. The configuration and

attachments of in-plane system is shown in Figure 4-23. The shell element has similar

specifications to the five facade panels with normal height and width of 3600 mm and

1500 mm, respectively, per bay. Therefore, the weight of each shell element is equal

to that of five facade panels. Stiffness of each horizontal connection, which represents

an energy absorbing connection, is calibrated in order to incorporate them into the

total lateral stiffness of the whole structure. It should be noted that these shell

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elements have a similar behaviour to the shear or infill walls in order to combat the

seismic excitations. The vertical connections should be able to support the mass of the

facades and provide minimum deformation. Mass of the façade shell is computed as

113 kN based on all previous assumptions. Various values of stiffness for the vertical

connections were evaluated to have a good idea about the behaviour and performance

of the proposed system.

4.4.4.1 Results and Discussion

Responses of a plastically behaving structure with conventional facade system are

compared with those of the structure equipped with shell elements. Stiffness of the

horizontal springs which attach the shell elements to the main frame is changed

between practical values to evaluate the response of the main structure. In Figure 4-24

, Table 4-8 and Table 4-9 these comparisons presented.

Figure 4-24: Relative displacement of top level of main structure with different stiffness of bracket under 1940 El Centro Earthquake

Systems with fixed (fixed spring shell) or very high (1.0e + 9 kN/m) stiffness

represent infill walls or shear walls which dramatically decrease the peak lateral

displacement of structure and the corresponding RMS values. Comparison between

different stiffness of the horizontal connections demonstrated that the higher the

stiffness, the more the RMS would be.

-200

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Elastic structure with conventional façade Plastic structure with conventional façadePlastic K=10000 Plastic K=1000Plastic K=100 Plastic K=10

125

Table 4-8: Maximum lateral displacement of the structure in different value of shell spring stiffness in El Centro Earthquake

Structural System Maximum Lateral Displacement (mm) Elastic Conventional façade bracket 136.50

Pla

stic

S

truc

ture

( )

Conventional façade bracket 103.66 K=10,000 28.43K=1,000 78.13K=100 95.68K=10 98.43

Table 4-9: Root Means Square for in-plane-shell

Structural System RMS Elastic Conventional façade bracket 50.6

Pla

stic

S

truc

ture

( )


Since the distance between the shell panel and main structural frame is considered

to be 200 mm, then the in-plane displacement of the facade shell panels should be

evaluated as well. The movement should be controlled because of the practical

concern about this relative displacement of the panel and risk of its collision with the

main structural frame. Relative in-plane displacement of top floor facade panels with

different horizontal stiffness is shown in Figure 4-25

Figure 4-25: Relative displacement between shell and structure at top level under 1940 El Centro Earthquake

Top relative displacement with different values of shell spring stiffness is presented

in Table 4-10.

-25-20-15-10-505

10152025

0 3 6 9 12 15

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t (m

m)

Time (Sec)

Elastic structure with conventional façade Plastic structure with conventional façadePlastic K=10 Plastic K=100Plastic K=1000 Plastic K=10000

126

Table 4-10: Lateral Displacement of top façade panels under 1940 El Centro Earthquake

Structural System Maximum Lateral Displacement (mm) Elastic Conventional façade bracket 22

Pla

stic

S

truc

ture

( )

Conventional façade bracket 19 K=10,000 0.03K=1,000 0.28K=100 2K=10 16

Also, for the relative displacement of the spring shell facade with different

stiffness, the values of RMS are shown in Table 4-11.

Table 4-11: Root Mean Square of relative displacement of façade shell element at top level of structure

Structural System RMS Elastic Conventional façade bracket 7.9

Pla

stic

S

truc

ture

( )


It is seen that the proposed system is working very effectively in terms of reducing

the response of the primary structure. With the stiffness of around 1,000 , this

reduction reaches its maximum.

4.4.4.2 Conclusion

Based on the results of this section, it is demonstrated that incorporating the in-

plane stiffness of facade panels can be a feasible concept in order to reduce the

response of the primary structure. Determining the tensile strength of glass elements is

a crucial part in developing this idea. A stress-history interaction equation that is

useful for determining the tensile strength of glass with accuracy is the key parameter

for the proposal. As the facade panels are only located on the perimeter of the

structure, then the concept cannot be used for the internal frames of structures.

127

Moreover, the results are very sensitive to the energy content and dominant

frequencies of the applied earthquake; therefore, the possibility of both near- and far-

field earthquake motions should be evaluated in order to assess the capability of the

proposed system.

4.5 Finite Element Modelling Using ANSYS APDL

4.5.1 Introduction of Smart Bracket with Combined Shear and Axial

Movement

Fundamental vibration period and frequency of structures have direct correlation

with their overall height. This frequency is within an approximate range of 0.3 to 3.0

Hz for low- to mid-rise buildings. On the other hand, frequencies associated with most

earthquake records are between 0.2 and 5 Hz depending on many factors including sub

soil conditions (Figure 4-27). Therefore, resonant conditions, in which the frequency of

seismic force is similar to that of building structures, can most probably happen during

seismic activities. In the case of seismic activities, low- to mid-rise buildings may be

susceptible to excessive inter-storey drifts, large shear forces, and noticeable

accelerations.

A comprehensive assessment procedure for ensuring the feasibility of the proposed

system to mitigate seismic hazard is performed in this section of the thesis. In the

previous sections, it was shown that facade panels could be considered as structural

elements and as movable panels; then they have the potential to reduce buildings'

lateral displacement if axial stiffness of facade brackets is such that yielding can occur

during earthquake excitations. In this section, a new concept is introduced in order to

utilize dynamic capability of movable facade panels in order to reduce the response of

primary structure during different earthquake events. Adding more functions to the

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bracket elements in addition to carrying the self-weight of panels is novel in designing

the new generation of facade panels. Selection of suitable materials to provide the

desired behaviour would be a part of future works related to the proposed idea.

4.5.2 Development of Smart Passive Façade System-Assigning a Nonlinear

Behaviour to Façade Connection

Initial investigations have illustrated that the best results in terms of reducing

ductility demand on main structural components can be achieved using a facade

moving at the TMD frequency (bracket with low stiffness). On the other hand, facade

brackets should be designed to withstand severe winds; it means that they should

display reasonable vertical and horizontal displacement and stiffness. It should be

noted that, based on facade design calculations, each facade panel should have a

plus/minus reasonable displacement during the force application. The main concept is

to replace the conventional bracket elements, which have a rigid behaviour, with new

bracket elements with low shear and axial stiffness to yield and be sacrificed and

hence absorb energy in the case of extreme earthquake excitations

Figure 4-26: Proposed Multi-linear behaviour of the façade bracket acting as axial damper system

For

ce (

n)

Deformation (mm)

129

They should have the ability to deform under tensile stress. The advanced connectors

are modelled as a nonlinear translational spring element, which could incorporate a

bilinear behaviour with strain hardening and inelastic unloading. The proposed

behaviour of bracket elements is shown in Figure 4-26. The excessive movement of

smart facade panels are controlled by the third slope. The last slope which has much

higher stiffness acts as a brake against the panels back and forth movement. The

required stiffness is calculated based on the maximum allowable movement of facade

panels in practice. Third phase of force-deformation behaviour of the panel is

associated with a higher stiffness ratio to ensure façade remains at a safe distance from

the main structure. The initial stiffness of these systems is designed based on the wind

loads for each specific region. The secondary stiffness for each individual facade

panel should be designed to allow yielding of the brackets after a threshold seismic

acceleration is reached.

For example, for a region which is located in an earthquake-prone zone with low

wind forces, the first stiffness is designed with low stiffness to increase the efficiency

of the system. For the region which is an earthquake-prone zone with high wind

forces, the first stiffness should be designed with higher stiffness to withstand

expected wind forces and the second slope of bracket stiffness should be designed to

allow yielding for mitigating the effect of earthquake. The second slope should be

designed specifically based on earthquake forces capable of yielding and forming

plastic hinges in the brackets. If the resulting facade displacements reach a particular

threshold value the stiffness is increased the stop further façade displacements.

130

4.5.3 Earthquake Records and Their Features

A 500 year return period (RP) event in high seismicity regions is considered for all

the seismic evaluations in this section. All records are normalised to 0.7g Ground Peak

Acceleration (GPA) which represents a high intensity earthquake.

Figure 4-27: Seismic hazard versus return period (Paulay 1992)

Two earthquake records with different frequency contents and durations are

selected to conduct a comprehensive evaluation on the performance of the proposed

system. Moreover, a series of Power Spectrum Density (PSD) evaluations are

performed on the time-history of the selected earthquakes, shown in Figure 4-28 and

Figure 4-29, respectively.

131

(a)Whole record (b) Comparision between first 10seconds with second 10 seconds

Figure 4-28: Displacement Power Spectrum Density for 1994 Northridge earthquake

(a)Whole record (b) Comparision between first 10seconds with second 10 seconds

Figure 4-29: Displacement Power Spectrum Density for 1963 Hachinohe earthquake

Displacement time-history of the records is applied to the structure using ANSYS

APDL program. It is important to note that the comprehensive evaluation of PSD is

necessary for designing and selecting the best stiffness of the brackets. For example,

in Northridge earthquake, the dominant frequency containing significant seismic

energy is evident during the first 10 sec of the record and the second 10 sec of the

0E+0

1E+4

2E+4

3E+4

4E+4

5E+4

6E+4

0 0.2 0.4 0.6 0.8 1

PSD

Dis

plac

emen

t 103

x [m

m^2

/Hz]

Frequency (Hz)

WholeRecord

0E+0

1E+4

2E+4

3E+4

4E+4

5E+4

6E+4

7E+4

0 0.2 0.4 0.6 0.8 1

PSD

Dis

plac

emen

t 103

x [m

m^2

/Hz]

Frequency (Hz)

First 10 seconds

Second 10 seconds

0E+0

5E+3

1E+4

2E+4

2E+4

3E+4

3E+4

0 0.2 0.4 0.6 0.8 1

PSD

Dis

plac

emen

t 103

x [m

m^2

/Hz]

Frequency (Hz)

Whole Record

0E+0

5E+3

1E+4

2E+4

2E+4

3E+4

3E+4

0 0.2 0.4 0.6 0.8 1

PSD

Dis

plac

emen

t 103

x [m

m^2

/Hz]

Frequency (Hz)

First 10 seconds

Second 10 seconds

132

record does not contain high energy in the considered frequency range. For

Hachinohe record, it is seen that frequency content is almost steady during the whole

record.

4.5.4 Structural Models

In a finite element analysis, selection of mesh size and layout is critical. It is

usually desirable to use as many elements as possible in the analysis to improve

accuracy. However, such an analysis will require excessive computer modelling time.

In this analysis, adequate number of elements are chosen for both frames and facades

in order to obtain sufficient accuracy of the results without excessive use of computer

time after carrying out a convergence study. To show the feasibility of the concept,

ANSYS APDL package is used to model the external panel of a 3D 12-storey structure

with rigid and sacrificial brackets. Length of each bay is 8.0 meters and height of every

floor is considered to be 4.0 meters with the overall height of the structure of 48m. All

structural joints are modelled as a semi-rigid connection based on Euro code

(moment-rotation relation) to accurately model the actual behaviour of the structure..

If designed properly, ductility ratio of column and beam elements should be around 1-

1.5 and 6-8, respectively, to reach the overall ductility of 3-4 for the structure.

4.5.5 Bracket Element Behaviour

It has been proven that facade panels with particular weight and dynamic

characteristics can mitigate the response of the main structure. As mentioned earlier,

direction of the applied earthquake force could be horizontal or vertical; therefore,

bracket facade elements can be loaded in axial and/or shear directions. The key

criterion for designing the bracket stiffness is to have comprehensive knowledge about

the nature of earthquake forces. Then, a series of power spectral density (PSD)

133

analyses were conducted to evaluate the content and frequency range of the selected

records. Facade brackets are tuned to the main frequency of structure to maximise

deformation and hence the load on the brackets. Brackets' axial stiffness should be

designed to withstand at least the design wind load with negligible facade movements.

It is assumed that each panel is attached to the slab of main structure with four

brackets, one at each corner. The value of the calculated stiffness is higher than the

value calculated based on considering the facade panels as a multi Tuned Mass

Damper (TMD) along the height of the structure. It should be noted that the axial

force in the conventional brackets due to earthquake excitation would be another

parameter, which should be considered in calculations.

Figure 4-30: Configuration of the proposed damper system

COMBIN14 element is used in ANSYS for modelling a simple linear spring-damper

element, which represents the assumed behaviour of the bracket element. Shear

stiffness of the brackets is changed between some preselected values (100, 80, 50, 20,

15, 12, 10, 8, and 5 N/mm) to gain an understanding of their respective effects and the

results are shown in the following figures. The proposed damper system, which

replaces the conventional bracket system, principally behaves like a Viscoelastic

damper which is shown in Figure 4-30. The conventional bracket system has

dimensions of length of 350mm, width of 280mm and thickness of 20mm. other

technical specification of the conventional bracket is listed as below:

ElasticModulusofeffectivesection:4

28000

Areaofeffectivesection: ∗ 5600

Allowableflexuralstrenght:1.65

4.09 . .

It consists of two plates which are filled with a viscoelastic material with a high

value of axial stiffness in one direction and low value of shear stiffness in the

perpendicular direction. One end of the bracket system is connected to façade element

and the other end to the main structural element (column or slab). The inner plate is a

separator between the two layers of visco-elastic material in order to allow more

relative displacement to the bracket system.

Conventional bracket system which represent rigid bracket element has length of

350mm, width of 280mm and thickness of 20mm. This bracket element

4.5.6 Criteria for Evaluation of the System

To gain an insight into the performance of the system, three criteria are evaluated

and the results are shown below. Top lateral displacement and acceleration of the

primary structure, storey drifts, and root mean square of displacement are the criteria

for interpreting the overall performance of the system.

4.5.6.1 Lateral Displacement Control

In this research, the effects of implementing novel bracket facade elements with low

shear stiffness, which can be installed in a 12-storey structure system, are evaluated.

The conventional bracket elements are replaced with these novel connections, which

are installed between the main structure and glazing system on all four faces of the

building. One outer frame, along the same direction as the earthquake excitation, is

134

135

evaluated and the results are shown below. Time-history responses of top lateral

displacement of the structure under different shear stiffness and different bracket

placements under both Northridge and Hachinohe earthquake excitations, obtained

from analytical models using computer program ANSYS, are illustrated in Figure 4-31

and Figure 4-32.

Figure 4-31: Relative displacement between top and bottom of the primary structure during Northridge record

It is observed from these figures that the novel facade system can reduce the top

displacement of the structural model during most parts of the time-history of the

earthquake. For Northridge record, it can be seen in Figure 4-31 as well as Table 4-12

that, with the bracket, with low shear stiffness, one adequate reduction in lateral

displacement under earthquake excitation can be achieved. Notably, based on the

interpretation of PSD, larger responses for Northridge record take place in the first 20

sec of the excitation

-300-250-200-150-100-50

050

100150200250300

0 4 8 12 16 20

Lat

eral

dis

palc

emen

t (m

m)

Time (Sec)

Conventional k=100 k=20 k=10 k=1

136

Figure 4-32: Relative displacement between top and bottom of the primary structure during Hachinohe record

Bracket elements with the shear stiffness of 10 N/mm are tuned to the frequency of

primary structure which has a similar frequency to the dominant frequency of

Northridge excitation. The maximum reduction of the top lateral displacement of the

structure subjected to Northridge excitation takes place during the first 20 sec of

excitation. According to these results, it is shown that, with the use of low shear

stiffness brackets, the overall lateral displacement of the primary structure subjected to

seismic load can be decreased.

Table 4-12: Maximum top lateral displacement in primary structure incorporating low shear stiffness bracket facades during Northridge record

Northridge Record Maximum Displacement (mm) % Reduction

Conventional 205 -

She

ar S

tiff

ness

(N

/mm

)

100 214 - 80 211 - 50 214 - 20 168 18 15 161 22 12 148 28 10 136 35 5 124 40 1 139 32

It is also demonstrated that values of 1 and 20 N/mm have less effect on response

reduction compared to the values of 5 and 10 N/mm. To be more precise, it is

concluded that the optimum range is from 5 to 15 N/mm for the Northridge

earthquake record. For Hachinohe record, it can be seen in Figure 4-32 as well as

-300-250-200-150-100-50

050

100150200250300

0 4 8 12 16 20

Lat

eral

dis

palc

emen

t (m

m)

Time (Sec)


137

Table 4-13 that with the bracket with low shear stiffness adequate reduction of lateral

displacement under earthquake excitation can be achieved. Bracket elements with the

shear stiffness of 20 N/mm are tuned to the frequency of the primary structure which

has a similar frequency to the dominant frequency of Northridge excitation.

Table 4-13: Maximum top lateral displacement in primary structure incorporating with low shear stiffness bracket facades during Hachinohe record

Northridge Record Maximum

Displacement (mm) % Reduction

Conventional 126 -

She

ar S

tiff

ness

(N

/mm

)

100 108 1480 94 2550 79 3720 66 4715 71 4312 78 3810 83 345 79 371 75 40

According to these results, it is shown that, with the use of low shear stiffness brackets,

the overall lateral displacement of the primary structure subjected to Hachinohe

seismic load is decreased. It is also demonstrated that values of 1 to 20 N/mm have

similar effects on response reduction compared to the value of 20 N/mm which has

maximum reduction. To be more precise, it is concluded that the optimum range is

from 10 to 20 N/mm for the Hachinohe earthquake record.

4.5.6.2 Drift Control

The in-plane seismic assessment of glass facade systems requires an estimation of

the likely in-plane drift demand from the earthquake action. AS 1170.4 (2007), clauses

5.4.4 and 5.5.4, specify that, “the inter-storey drift in the ultimate limit state, calculated

from the forces determined according to strength and stability provisions, shall not

exceed 1.5% of the storey height for each level” and “the attachment of cladding and

138

facade panels to the seismic-force-resisting system shall have sufficient deformation

and rotational capacity”. Therefore, for the typical floor height of 3600 mm, the

maximum allowable relative storey deflection is 54 mm. An earthquake-induced drift

is a controversial issue for the safety assessment of building structures. Seismic drift

demand on buildings can be investigated in many ways using elastic or inelastic

approaches with static or dynamic analyses (Sivanerupan et al. 2008). Drift levels are

set to ensure that the design storey drift does not exceed that level, which is consistent

with the ductility of the available element based on structural detail requirements of

AS1170.4. In-plane and out-of-plane seismic drift demands of facade systems should

be estimated if low shear stiffness is used in the bracket elements. The drift capacity of

a framed glazed system before glass breakage depends on the edge clearance and

aspect ratio. These items should be considered in detail to evaluate in-plane and out-of-

plane deformations of the frame. If the drift between the facade panels is higher than the

limit, then the facade is considered unsafe and it is necessary to carry out further assessments.

The results indicate that framed glazed facade systems with typical minimum edge

clearance in regular buildings are not vulnerable to the selected earthquakes. In

computer modelling, the facade panels are assumed to be conventional curtain wall

systems, which are attached to the slab of the main structure at four points. Point fixed

(frameless) glazing is another type, which needs separate analytical modelling and

rational testing to assess the contribution to response reduction. Commercially

available curtain walls provide serviceability movements in the range of 10-15mm

(30% drift) for inter-storey displacement demand. These systems seem to reach this

ultimate condition at the inter-storey drift of approximately 30-40 mm. Drifts for each

storey for primary structure with different stiffness for shear bracket façade elements

139

during Northridge and Hachinohe earthquakes are shown in Figure 4-33 and Figure

4-34.

Figure 4-33: Drift for primary structure with different stiffness for shear bracket façade elements during Northridge earthquake

As illustrated in Figure 4-33 drift for the floors with the shear stiffness of 10 N/mm

for bracket elements is decreased by the maximum value of 44% which is a promising

result. Values of in-plane drifts of the primary structure for the selected shear stiffness

are shown in Table 4-14 for Northridge earthquake.

Table 4-14: In plane drift of primary structure with different bracket stiffness during 1994 Northridge earthquake (in mm)

0

10

20

30

40

50

0 1 2 3 4 5 6 7 8 9 10 11 12

Dri

ft b

etw

een

stor

eis(

mm

)

Storey


Floors Conventional 100 20 10 1

12 11.0 10.1 10.0 9.8 10.111 11.0 10.5 8.6 11.0 9.010 15.6 14.9 11.8 10.9 11.09 19.8 17.9 14.3 11.5 13.08 22.7 19.8 16.2 12.7 14.47 23.8 22.1 18.4 14.3 15.96 23.6 23.9 19.6 14.9 16.75 24.9 24.8 19.8 14.5 16.34 26.5 24.7 21.1 15.6 16.73 28.9 24.7 22.8 17.3 16.92 26.7 22.2 21.1 16.2 15.41 14.2 11.7 11.1 8.6 8.2

140

Figure 4-34: Drift for primary structure with different stiffness for shear bracket façade elements during Hachinohe earthquake

As illustrated in Figure 4-34, drift for the floors with the shear stiffness of 20

N/mm for bracket elements is decreased by the maximum value of 60% which is

another promising result. Values of in-plane drifts of primary structure for the selected

shear stiffness are shown in Table 4-15 for Hachinohe earthquakes.

Table 4-15: In plane drift of primary structure in case of different bracket stiffness during 1963 Hachinohe earthquake (in mm)

4.5.6.3 Acceleration of Primary Structure

Controlling the acceleration of structures is very crucial during earthquake

excitations. Excessive acceleration in a structure can cause severe damage to sensitive

equipment. Top floor acceleration of the primary structure is compared using various

bracket facade stiffness values and results are shown for both earthquake records. As

0

10

20

30

40

50

0 1 2 3 4 5 6 7 8 9 10 11 12

Dri

ft b

etw

een

stor

eis(

mm

)

Storey


Floors Conventional 100 20 10 1

12 24.8 21.2 11.3 17.9 16.711 24.7 21.0 12.2 19.4 18.010 21.6 18.4 11.9 18.9 17.69 20.8 17.0 9.5 15.2 14.18 17.4 14.2 8.4 13.4 12.47 20.1 16.2 7.6 12.2 12.06 22.9 19.0 8.9 13.2 12.15 24.0 19.9 9.9 15.0 14.34 23.5 19.5 10.1 14.00 13.03 24.7 20.9 11.1 17.6 16.32 26.2 22.2 15.7 18.10 19.91 28.1 23.0 17.8 20.00 20.2

141

illustrated in Figure 4-35and Figure 4-36, using low shear stiffness for facade bracket

elements leads to reduction in overall acceleration on the top floor of the primary

structure by a reasonable value during both excitations.

Figure 4-35: Acceleration in top floor of primary structure with different shear stiffness for bracket facades during Northridge earthquake

Figure 4-36: Acceleration in top floor of primary structure with different shear stiffness for bracket facades during Hachinohe earthquake

4.5.6.4 Root Mean Square of Top Displacement

Root Mean Square (RMS) is a statistical measure of the magnitude of a varying

quantity. To have a better view of the façade performance, RMS of top displacement

using different bracket stiffness is compared in Table 4-16 and Table 4-17. As can be

seen, using shear stiffness of 10N/mm for Northridge earthquake and 20N/mm for

Hachinohe earthquake, the RMS values are almost halved compared to the system

with fixed traditional facade panels.

-8000

-6000

-4000

-2000

0

2000

4000

6000

8000

0 4 8 12 16 20

Acc

eler

atio

n (m

m/s

2 )

Time (Sec)


-8000

-6000

-4000

-2000

0

2000

4000

6000

8000

0 4 8 12 16 20

Acc

eler

atio

n (m

m/s

2 )

Time (Sec)


142

Table 4-16: Root mean square of top displacement using different value of shear stiffness during 1994 Northridge earthquake

Type of Structure RMS, mm % Reduction Conventional 67.2 -

100 69.2 -80 70.1 -50 73.1 -20 47.2 3010 31.9 531 42.2 37

Table 4-17: Root mean square of top displacement using different value of shear stiffness during 1963 Hachinohe earthquake

Type of Structure RMS, mm % Reduction Conventional 62.5 -

100 55.3 1180 51.3 1850 47.3 2420 24.5 6010 42.1 321 38.3 38

4.5.7 Conclusions

From the numerical results of the current study, it is found that bracket facade

elements with low shear stiffness can mitigate the response of structures. Reaching a

lower value of stiffness would not lead to excessive movement of panels, as was

proved for lower axial stiffness. In this study, the most efficient configuration of shear

brackets is their placement at the top 1/3 height of the building. However, it must be

emphasized that, to reach the maximum performance, using designed brackets on all

floors of the structure is recommended; but, this adds additional cost to construction.

All results are based on using a simple linear element with low shear stiffness and

higher axial stiffness. Selection of bracket material is a crucial aspect in terms of the

practicality of the proposed idea. Viscoelastic material, code ISD-111 VE polymer,

manufactured by 3M Company would be one possible choice for material selection

and bracket experimental testing in future works. At this stage of computer analysis, a

143

new concept is introduced to utilize dynamic capability of movable facade panels to

reduce the response of primary structure during different earthquake events. Adding

more functions to the bracket elements in addition to carrying self-weight of panels is

novel in designing a new generation of facade panels. Selection of suitable materials

for providing the desired behaviour would be a part of the future work on the proposed

idea. Maximum in-plane and out-of-plane drift capacity of the proposed facade panels

are evaluated to demonstrate the overall feasibility of the system.

In-plane drift capacity of glazed facade systems should be considered as a key

factor for designing the bracket stiffness. Designing an advanced facade connection is

a complex process and will be time-consuming if used for commercial purpose. But, it

could be justified by the overall increased efficiency of structural facade functioning

in regard to dynamic response, such as a seismic situation. The advanced facade

connection can provide better uniformly distributed energy dissipation over the height

of the building without involving any structural members. The experimental testing

under the conditions close to the realistic physical structure should be conducted. The

main idea behind the experimental phase is to validate the numerical model as well as

observing the real behaviour of the system subjected to a number of earthquake

records.

144

CHAPTER 5

MID-RISE 10-STOREY

STRUCTRAL MODELS

145

5.1 Introduction

In this chapter, the performance of the proposed movable façade system models,

described in Chapter 4, is investigated, comprehensively, with modal analyses and

nonlinear dynamic time-history analyses. The overall behaviour of the mid-rise

structural models with movable dissipative cladding is compared to the same structure

with conventional façade system, and the dynamic effects and behaviour of the new

cladding system are compared across the models with different axial and shear

stiffness connections. The effects of the movable cladding system on the seismic

response of the bare-frame structures are studied by performing modal and nonlinear

dynamic time-history analyses.

Analytical calculations have shown that the movement of cladding system

significantly influences the seismic response of mid-rise structural buildings with

movable dissipative cladding in moderate to severe earthquakes. Normally, fix-

cladding systems reduce the fundamental period of the structural building by few

percent. Effects of the movable cladding on the fundamental period of the structural

building are discussed in this section as well. Additionally, effects of the cladding on

inter-storey drifts, floors displacement, floors acceleration, plastic hinge rotations in

beam/column connections and base shear are going to be discussed in this chapter.

Time-history analyses of each model are performed with selected recorded ground

motions to determine the structural reliability of the proposed connections.

5.2 Modelling Approach and Assumptions

The modelling approach used for the mid-rise structural building with movable

claddings is discussed in this section. Finite element assumptions, building

dimensions, actual beam-column member sizes, equivalent section member sizes,

146

connection details, and total seismic masses are discussed in details in this chapter. It

should be mentioned that defining the reinforced concrete material with all details and

modelling of structural interaction between the steel bars and the concrete material is

beyond the scope of this computer modelling. Hence, a smeared model is used

combining the properties of concrete and steel in the complex structural model in the

3-D 10-storey structure.

5.2.1 Finite Element Analysis

The program selected for the numerical analysis is ANSYS APDL. This

programme is used for generating the geometry, boundary conditions and loading

conditions of the models for both analyses. To decrease the computational time and to

simplify the structural modelling, one-dimensional frame elements were selected for

beams and columns and two-dimensional plane stress elements were chosen for facade

panels in this program. The beams connecting the two columns of the equivalent bay

are modelled as elastic elements with rotational springs at both ends. Rotational

springs are utilised to model the equivalent strength and stiffness of all beam/column

connections. Each spring represents the cumulative strength of half the number of

simple connections.

5.2.1.1 Meshing Size

In a finite element analysis, selection of mesh size and layout is crucial. Usually, it

is ideal to use as many elements as possible in the analysis to improve accuracy.

However, such an analysis will require excessive computer time. In this analysis,

adequate numbers of mesh elements were selected for both structural frame and

façade components to obtain sufficient accuracy of results without excessive

computational times.

147

5.2.1.1.1 Column, Beam and Slab

The mesh discretization must balance the need for fine mesh to give an accurate

stress distribution and reasonable analysis time, so the beams and columns are divided

into three sections along their length for meshing in all numerical modelling.

5.2.1.1.2 Bracket (Spring Beam)

For bracket façade, the optimal solution is to use a fine mesh in areas of high stress

and a coarser mesh in the remaining areas. However, this kind of meshing would

definitely increase calculation time. Each Bracket element, 50cm in length, is divided

into five equivalent sections for meshing in all numerical modelling.

5.2.1.1.3 Façade Column

As the façade column is modelled as a linear beam element, then mesh size of

facade column is assumed to be 10cm in all numerical models.

5.2.1.2 Direction of Applied Loads and Building Boundary Conditions

The chosen earthquake record acceleration was applied in two directions (X and Z

directions) in the 3-D structural models, at the base of the structure, as shown in

Figure 5-1. The supports at the base of the structure were modelled as a rigid joint,

restrained against translation and rotation in x, y and z directions. The vertical gravity

loading on the structure was in the form lumped masses applied to the beams.

148

Figure 5-1: Direction of applied earthquake in the 3-D model

5.2.2 Element Behaviour and Structural Modelling

In this section, as the behaviour of proposed façade system is evaluated in mid-rise

structures, then three-dimensional (3-D) structural models is going to be modelled.

Sectional dimensions which are used for the model are listed in Table 5-1and more

details are given in Appendix-A.

Table 5-1: Structural sections for beam and column elements

Moment-resisting frame Storey/Floor column (mm) Beam (mm)

10 400X400 400X5009 400X400 400X6008 400X400 400X6007 400X400 400X6006 500X500 400X6005 550X550 400X6004 600X600 400X6003 650X650 400X6002 700X700 400X6001 700X700 400X600

Floors and total building heights are assumed 3.6 meters for each level and 36

meters, respectively, for the 3-D structure models.

Applied Earthquakes

Y

X

Z

149

Figure 5-2: Front view of exterior elevation of the 3D frame model

The view of frame structure used for the 3D models in this section is shown in

Figure 5-2. The model has four and three bays in X and four bays in Z directions,

correspondingly. The length of each bay is considered as 6 meters for both directions.

It is assumed that building mass is distributed equally at each floor along the height of

the structure. Figure 5-3 shows beams and columns layout in building structure plan.

Figure 5-3: Plan view of the 3D frame model (dimensions are in mm)

150

5.2.2.1 Material Properties

The concrete materials selected for the study are listed in Table 5-2. Equivalent

sections with equivalent material properties were used for numerical modelling of the

main structure, as reinforced concrete material is hard to model in ANSYS. The

properties used for the equivalent material are the same as concrete except for

modulus of elasticity.

Table 5-2: Selected concrete properties

Criteria ValueCompressive strength, f'c (MPa) 32

Young’s modulus, E (MPa) 30,000Density(kg/m3) 2400

Poisson’s ratio, � 0.2

5.2.2.2 Consideration of Seismic Mass

In the analysis of the structural models, the mass of the building is assigned to the

beam-column nodes. This representation simplifies the assignment of mass to the

frame while precisely represents the distribution of mass throughout the structure. The

mass values account for the framing, floor deck and roof, ceilings/flooring,

mechanical/electrical and interior partitions. Masses are defined on middle of beam

elements (middle of each bay) as lumped mass and their distribution is according to

Table 5.3. Mass element was defined in all three directions in each floor to represent

slab mass in numerical modelling of the 3D model.

Table 5-3: Mass values for bare frame model

Storey Mass value at each beam-column node in all stories (N.sec2/m)

1-10 80

151

5.2.2.3 Design of Structural Elements

As mentioned, modelling of reinforced concrete elements consisting of concrete

and steel rebar is complex because the interaction between these two materials needs

to be modelled precisely with all details. Such detailed modelling, for 10 or 30-storey

structure models including façade elements, is almost impossible to converge.

Therefore, analysis and design of building structural sections were performed using

SAP2000 program. The design earthquake loads are based on the provisions of the

Australian Standard 1170.4, and all relevant assumptions are shown in Table 5-4.

Table 5-4: Assumed factors for earthquake design

Factors Assumed value Importance level of structure (BCA) 3

Annual Probability of exceedance (P) 1/1000 Probability factor (Hemalatha and Jaya) 1.0

Hazard Factor(z) 0.08 Site Sub-Soil Class Be (Rock)

Earthquake Design Category(EDC) III

5.2.2.3.1 Equivalent Sections for Structural Beams and Columns

Equivalent sections are used in order to simplify the ANSYS finite element models.

Calculation of an equivalent section is very simple and can be found in many concrete

design handbooks, but for the same of completeness, it is included below in Figure

5-4. Section details and their equivalent section for each type of structural elements

are shown in Appendix A.

152

Gross Section Transformed Section Equivalent Section

Figure 5-4: Reinforced concrete section and assumed equivalent section

5.2.2.4 Assumption of Strong Column and Weak Beam Connection by

Semi-Rigid Connections

During the earthquake episodes, the buildings undergo large lateral displacements

and plastic hinges develop most likely at the beam member ends. Semi-rigid

connections need to be modelled in 3D models in order to have results that ensure this.

Therefore, the proposed connection in Figure 5-5 is used to represent an ideal model.

Four triangular plates, which were modelled by connecting groups of three nodes,

connect the columns to four corners of the slab. These four rectangular plate elements

around the boundary of the slab were created to represent the drop panels of the

building. These elements were used to provide a semi-rigid connection between the

columns and the beam/slabs.

kd

d-kd

b

d

As2

As1

nAs

N

A

153

Figure 5-5: Details of proposed column/beam connection

The beam-column stiffness and resisting forces are transformed from local

coordinates to the global coordinates using the correlational transformation, which

captures secondary effects from large displacements and P-Δ effects. Figure 5-6 shows

typical force-deformation hysteretic curve for modelling plastic hinges in reinforced

concrete beams.

Figure 5-6: Modelled force-deformation hysteretic curve for modelling plastic hinges in reinforced concrete beams

5.2.3 Details of Cladding System (Double Skin) Modelling

Double-Skin Facade (DSF') or "airflow" facade is assumed to be used in the

structural models. As compared to conventional facade systems, DSF's can reduce

Columns

Slab location

Equivalent Beam

154

energy consumption by 30%. They can provide natural ventilation even in structural

buildings, and providing valuable noise reduction. They also create a visually

transparent architecture that is impossible with conventional curtain wall facades with

similar thermal properties.

5.2.3.1 Material Properties

Generally, glass types are selected depending on their location in the building but

in this research, all glasses are assumed to have same the dimensions and material

properties; and dimensions of window panes are 180 cm in height and 150cm wide. It

is assumed that the insulating glass unit (IGU) panes consist of two 6mm glass panes

with a spacer of 12mm in diameter. It should be noted that 25 mm IGUs are typically

used where safety is not a concern, and heat-strengthened IGUs are used when the

panes are located within 45cm of the ground or within 120cm of a doorway. Material

properties used for numerical modelling are listed in Table 5-5.

Table 5-5: Material properties of façade panel components

Glass (Simax glass)

Rubber Aluminium

Young’s modulus (E) MPa 64,000 0.7 70,000Tensile strength (MPa) 35-100 60 240 Tensile yield strength 33 16 300

Allowable tensile stress (MPa) 3.5 5.5 386 Compressive yield strength (MPa) 150 30 530

Poisson’s ratio 0.2 0.45 0.33 Density (kg/m3) 2230 1100 2700

5.2.3.2 Façade Column Modelling

In order to investigate the effect of the movable cladding on the building response,

accurate finite element models of the cladding systems need to be developed. In order

to have more accurate results and to model the façade panels closer to reality, the

stack joint which connects two consecutive façade columns need to be defined and

155

modelled. The panels are modelled as BEAM188 elements with specific material

properties to represent four of the façade panels in reality. In order to define the joint

in ANSYS, one small element was defined with a length of 100mm. All three

displacements in the X, Y and Z directions are fully restrained at both ends of the

element, but rotations at both ends are allowed which is exactly how an ideal joint

behaves. The façade load bearing elements are assumed rigid, and they are modelled

with two-dimensional frames comprising rigid elastic beam elements. Figure 5-7

illustrates how successive façade panels are modelled as one single beam element in

each storey in ANSYS modelling. It should be noted that, the façade column elements

represent four façade panels in reality in terms of weight and dynamic response

(frequency of movement) in all modelling. Claddings consist of full storey-height

panels and are attached with horizontal bearing connection and flexible lateral

connection to the slab structure (Figure 5-8).

156

(a) Top view (b)Front view (c) Side view

Figure 5-7: Schematic view of facade column element and their configuration in each floor

The bearing connections resist the gravity loads of the panel and flexible lateral

connection allows a controlled lateral deformation between the frame and the outer

layer of cladding panels. The flexible connections are modelled as nonlinear zero-

length spring elements in ANSYS program and have very low modulus of elasticity in

direction perpendicular to applied earthquake direction.

Tributary area of structural column

Equivalent column element


Bracket (shear/axial Connector)

Structural column element

Gap between two skins

Panel 4 Panel 3 Panel 1 Panel 2


Structural column element

Bracket

157

Figure 5-8: Elevation view of façade connection

It should be mentioned that the rotational capacity of the connections is very small

and can be neglected. Since the bearing connections are very stiff in the horizontal

direction, the horizontal displacement, and rotation of the middle of the panel are

approximately equal to the displacement and rotation of the mid-span of the floor

beam. Therefore, the horizontal bearing connection is transformed into a displacement

constraint in the middle of the panel that Figure 5-8 panel’s displacements to the floor

beam displacements at mid-span. The detailed drawings of the designed bearing

connection and flexible lateral connection are shown in Section 6 in this chapter.

5.2.3.3 Details of Bracket Modelling

In primary numerical modelling, the bracket elements were modelled using

COMBIN element, a spring-damper element, but, in order to have more accurate

results, these elements had been replaced by BEAM188 elements in detailed

numerical modelling. Material properties of the BEAM188 elements were defined in a

Roller support Bottom Supported panel

Top hung panel 2600

90

0 10

0

158

way that they represent and behave in the same way as a COMBIN element in bracket

modelling (axial and shear behaviour). Therefore, literally the bracket elements are a

beam element with spring behaviour. The “rigid” cladding system represents common

systems used in regular multistorey buildings in modern construction. The force-

deformation curve of the axial connection which is for a hyper-elastic material is

shown in

Figure 5-9.

Figure 5-9: Defined Force-deformation curve of the axial connection

Shear brackets consist of two connectors; one is called gravity connection which

behaves like a rigid link and attach the outer skin of the façade system to the structural

slab and the other one is the damper connection which has a low stiffness in axial

direction. The force-deformation curve of the “shear” connection, which is the

modified version of axial connections, is shown in Figure 5-10. This connection is

modelled in a way that provides a rigid connection in compression but with much less

force resistance in lateral direction. The shear bracket consists of a single linear line

instead of multi linear stiffness.

F(N

)

d (mm)

500-1000 N/mm

20 80 100

10-20 N/mm

I II III

K1

K2 K3

159

Figure 5-10: Defined Force-deformation curve of the shear connection

In building structure, it is assumed that the façade panels are attached to the main structure

on all four sides. When applied earthquake is in X direction, bracket façade elements on Side

1 and side 3 move in direction to yellow arrows and most of the bracket forces are axially

induced. Bracket façade elements on side 2 and side 4 move in direction of red arrows and

most of the bracket forces are shear induced. Respectively, if earthquake applies in Y

direction, then behaviour of brackets on sides 1 and 3 will be changed to shear and sides 2 and

4 to axial. The proposed concept is shown in Figure 5-11.

Figure 5-11: Plan view of damper connections to the main structure and their behaviour in applied earthquake

K d (mm)

F (

N)

Linear Stiffness

Side 4

Sid

e 3

Side 2

Sid

e 1

Applied load

X

Y

160

5.3 Types of Numerical Modellings

5.3.1 Modal Analysis

Analysis of a building’s mode shapes gives an indication of the deformed shape of

the building during an earthquake. Depending on the frequency content of the ground

motion, one or more of the modes can be excited, and the deformed shape will then be

a superposition of the participating modes. A modal analysis was completed to

evaluate the effect of the cladding panels and their movable connections on the modal

periods and mode shapes of the 10-storey building. An analysis of a building’s

vibration periods provides information on how the building might respond to lateral

excitation. Depending on the predominant frequency of the ground motion, a building

with a shorter (or longer) fundamental period might be subjected to higher (or lower)

intensity shaking.

5.3.1.1 Results

In this section, using more realistic modelling approaches, the fixed cladding

system does not significantly affect the vibration characteristics of the building. There

is only 3.5% difference between the fundamental period of the bare frame and system

with rigid bracket cladding. The first three vibration periods are given in Table 5-6 for

the 3D models. The first mode shape of all models is almost linear, and the modal

ordinates are nearly identical.

Table 5-6: Modal vibration periods of models with bare-frame, fixed and flexible facades

Model Period (seconds)

Model 1 Mode 2 Mode 3 3D 3D 3D

Bare-frame 1.020.98 1.04

0.32 0.30 0.36

0.18 0.17 0.15

Fix facade Flexible facade

161

The effective modal mass percentage is a reliable measure on how much of the

total mass of the structure is participating in each vibration mode. The effective modal

mass percentages for the first three modes are provided in Table 5-7. The higher the

effective modal mass, the more dominant the lateral response of the building for an

arbitrary ground motion. However, there may be specific ground motions (near fault

motions), building sites (directivity effects, soft soil), or structural features (tall

buildings) where second or third mode are also participating just as much or possibly

more than the fundamental mode.

Table 5-7: Participating modal mass percentages

Model Participating modal mass percentages (%)

Model 1 Mode 2 Mode 3

Bare-frame 86.087.1 79.1

9.7 8.3

14.0

3.3 1.6 5.6

Fix facade Flexible facade

The first three modes contribute to approximately 98-99% of the total lateral

response in all the analytical models in this study. This finding indicates that these

three modes capture essentially the entire lateral dynamic response of the building.

Figure 5-12: Comparison of effective modal mass percentages for the first three modes

The first mode contributes over 80% to the total response, which is typical of well-

designed concrete frame buildings with an approximately uniform distribution of

86 87.181.7

9.7 8.314

3.3 1.6 3.6

0

20

40

60

80

100

Bare Frame Fix Façade Frame Flexible façade

Par

tici

pati

ng E

ffec

tive

Mod

al

Mas

s

Mode

Mode 1

Mode 2

Mode 3

162

structural properties and mass. Comparison of participating modal mass percentages

for the first three modes of the 3D structure is shown in Figure 5-12.

5.3.2 Dynamic Time-History Analysis (Holistic Nonlinear Time-History

Analysis)

The time-history analyses give better insight into how the building and cladding

system respond to the applied earthquake ground motions. The consequence of

nonlinear yielding, energy dissipation and damping of the cladding system are better

understood through time-history analyses. The seismic loadings applied to the

structural models in this study were from existing past earthquake records. These

earthquake records are time histories of horizontal ground accelerations. Earthquakes

have various properties such as duration of strong motion, the range of dominant

frequencies and peak ground acceleration (PGA). For that reason, they would have

different effects on the structures. The range of dominant frequencies as well as the

duration of the strong motions were kept similar to achieve a consistent comparison of

the response of a structural model to different earthquakes. It should be mentioned that

dynamic time-history analyses were performed to further investigate the seismic

behaviour of both the structural frame and the cladding connectors. The analyses were

performed using the Newmark integration scheme and 2% Rayleigh damping was

assumed for all hazard levels.

5.3.2.1 Dynamic Interpretation of Energy Content the Selected

Earthquakes (Power Spectral Density)

A 500 year return period (RP) event in high seismicity region is considered for all

seismic evaluations in this section. Four earthquake records as shown in Table 4.8

with different frequency content and duration have been selected to conduct a

comprehensive evaluation of performance of the proposed system.

Figure 5-13: Seismic and wind hazard versus excitation frequency (or period) (Paulay 1992)

Figure 5-13 shows that most of the earthquake energy lies between 1 and 10 Hz

which coincide with natural frequencies of most mid-rise building structures and

hence increasing the chance of resonant condition in mid-rise structures during

earthquake exaction. Moreover, a series of response Power Spectral Density (PSD)

evaluations have been performed using acceleration time-history of the earthquakes

and the results are shown in Figures 5-14 to Figure 5-17. Earthquake records are

normalized to 0.3-0.7g GPA that represents high intensity earthquakes.

Table 5-8: Characteristics of selected earthquake records

Earthquake Record Duration of strong motion

(seconds) Range of dominant frequencies (Hz)

Northridge(1994) 3.5 - 8.0 0.14 -0.69 El Centro (1940) 1.5 - 5.5 0.09 - 0.89

Kobe (1995) 7.5 - 12.5 0.21 - 1.22 Hachinohe (1968) 2.2 - 6.0 0.16 - 0.80

It is important to note that evaluation of PSD is necessary for designing and

selecting best brackets stiffness for each specific earthquake. As shown in Figure 5-14

163

164

to Figure 5-17, frequency content of the records spread over a range of frequencies

during excitation. In order to have better understanding of excitations frequency

content, each case is evaluated and interpreted individually as below. The frequency

of façade columns is set up and tuned for each earthquake acceleration record in order

to get the best results.

(a) Whole record (b) Comparision between first 20 secondswith second 20 seconds

Figure 5-14: Displacement Power Spectral Density for 1994 Northridge earthquake

In Northridge earthquake, the dominant frequency that contains significant seismic

energy is concentrated in the first 20 seconds of the record and the next 20 seconds of

the record does not contain much energy. Therefore, bracket stiffness design needs to

be set and tuned according to dominant frequency of Northridge earthquake that

happens in the first half part of the excitation. The maximum value of PSD for

Northridge record is around 50,000 mm2/Hz at frequency of around 0.12 Hz.

0E+0

1E+4

2E+4

3E+4

4E+4

5E+4

6E+4

0 0.2 0.4 0.6 0.8 1

PSD

Dis

plac

emen

t 103

x [m

m^2

/Hz]

Frequency (Hz)

Whole Record

0E+0

1E+4

2E+4

3E+4

4E+4

5E+4

6E+4

7E+4

0 0.2 0.4 0.6 0.8 1

PSD

Dis

plac

emen

t 103

x [m

m^2

/Hz]

Frequency (Hz)

First 20 seconds

Second 20 seconds

165

(a) Whole record (b) Comparison between first 20 secondswith second 20 seconds

Figure 5-15: Displacement Power Spectrum Density for 1940 El-Centro earthquake

The range of dominant frequencies in 1940 El-Centro record is between 0.09 and

0.69 Hz which is close to frequency of the 10-storey structural model. The dominant

frequency that contains significant seismic energy is concentrated in the first 20

seconds of the record, but the rest of the record also affects the dynamic behaviour of

the structural model. So, design of the bracket stiffness needs to be considered and

tuned based on the frequency contents of the whole record. The maximum value of

PSD for 1940 El Centro record is around 130,000 mm2/Hz at frequency of around 0.1

Hz. The range of dominant frequencies in 1995 Kobe record is wider in comparison to

the other two records and it is between 0.11 and 1.22 Hz. The dominant frequency that

contains significant seismic energy is concentrated in the first 20 seconds of the record

as before, and the rest of the record has little effect on dynamic behaviour of the

structural model. So, again the design of the bracket stiffness needs to be considered

and tuned based on dominant frequency, which is in the first 20 seconds of the record.

As shown in Figure 5-17 that frequency content is wide spread during the duration of

Hachinohe record.

00E+0

02E+4

04E+4

06E+4

08E+4

10E+4

12E+4

14E+4

0 0.2 0.4 0.6 0.8 1

PSD

Dis

plac

emen

t 103

x [m

m^2

/Hz]

Frequency (Hz)

Whole Record

00E+0

02E+4

04E+4

06E+4

08E+4

10E+4

12E+4

14E+4

16E+4

18E+4

20E+4

0 0.2 0.4 0.6 0.8 1

PSD

Dis

plac

emen

t 103

x [m

m^2

/Hz]

Frequency (Hz)

First 20 seconds

Second 20 seconds

166

(a) Whole record (b) Comparision between first 20 secondswith second 20 seconds

Figure 5-16: Displacement Power Spectrum Density for 1995 Kobe earthquake

The range of dominant frequency in Hachinohe record is between 0.16 and 0.8 Hz.

So, once again the design of the bracket stiffness need to be considered and tuned

based on frequencies covering the entire duration of Hachinohe earthquake.

(a) Whole record (b) Comparision between first part withsecond part of excitation

Figure 5-17: Displacement Power Spectrum Density for Hachinohe earthquake

5.3.2.2 Judgment on Engineering Demand Parameters

The global behaviour of the concrete moment resisting frame, due to earthquake

excitation, is typically described by maximum deformation, inter-storey drift ratios,

00E+0

05E+4

10E+4

15E+4

20E+4

25E+4

0 0.2 0.4 0.6 0.8 1

PSD

Dis

plac

emen

t 103

x [m

m^2

/Hz]

Frequency (Hz)

Whole Record

00E+0

05E+4

10E+4

15E+4

20E+4

25E+4

30E+4

35E+4

40E+4

0 0.2 0.4 0.6 0.8 1

PSD

Dis

plac

emen

t 103

x [m

m^2

/Hz]

Frequency (Hz)

First 20 seconds

Second 20 seconds

0E+0

5E+3

1E+4

2E+4

2E+4

3E+4

3E+4

0 0.2 0.4 0.6 0.8 1

PSD

Dis

plac

emen

t 103

x [m

m^2

/Hz]

Frequency (Hz)

Whole Record

0E+0

5E+3

1E+4

2E+4

2E+4

3E+4

3E+4

0 0.2 0.4 0.6 0.8 1

PSD

Dis

plac

emen

t 103

x [m

m^2

/Hz]

Frequency (Hz)

First 10 seconds

Second 13seconds

167

residual (permanent) drifts, floor accelerations, forces in bracket connections and base

shear force. These engineering demand parameters (EDPs) were determined during

the time history analyses of the analytical models and investigated in this section. In

this study, trends for maximum values of the engineering demand parameters, which

are typical of multi-storey concrete moment resisting frames, are discussed. It is better

to define a code for various stiffness configurations that are modelled in this study.

Building structures with the conventional bracket system; bracket system that has low

stiffness parallel to the applied earthquake and the bracket system that has low

stiffness perpendicular to the applied earthquake are called “Fixed”, “Axial” and

“Shear”, respectively. Moreover, the number which comes after each term represents

the value of the stiffness used in connection with the bracket system. For example,

“Shear-10” represents a bracket connection that is defined to move perpendicular to

the applied earthquake and has stiffness value of 10 N/mm.

5.3.2.2.1 Top Lateral Displacement

Effects of a particular earthquake on a building structure are usually evaluated by

maximum values of displacement at the top level of the building structure. The

effectiveness of the damper system was studied with various connections stiffness

during various earthquake records. Comprehensive response history of top-level

displacement of the three-dimensional 10- storey building structure is presented in this

section. Top floor displacement in case of “Rigid”, “Axial” and “Shear” connections

are extracted and compared in various figures and tables of this section. Comparison

of responses for the structure with rigid bracket facade and structure with axial bracket

facade showed that the proposed connectors were not able to reduce the peak values of

top floor displacement.

168

Figure 5-18: Time-history of top floor displacements of primary structure coupled with DSFs with different bracket connector stiffness during 1994 Northridge earthquake

But comparison of responses for the structure with rigid bracket facade and

structure with flexible shear bracket facade showed that the advanced connectors were

able to reduce peak values of top floor displacement. Shear-10 is chosen to compare

with rigid and axial bracket façade because it has the best performance and clearly

enhance the performance of proposed system. From Figure 5-18 it can be observed

that the maximum values of displacement occurred between short time intervals of

about 12 seconds up to 18 seconds. In addition, after approximately six seconds of the

1994 Northridge earthquake, the structure with flexible façade system began to

decrease response of the structure. The reduction is continued up to the end of the

record. The results for the same investigated parameter obtained by the frame structure

with the same connection properties in façade brackets under the 1940 El-Centro

earthquake excitation are presented in Figure 5-19.

-200

-150

-100

-50

0

50

100

150

200

0 5 10 15 20 25 30

Lat

eral

dis

palc

emen

t (m

m)

Time (Sec)

Conventional Axial Shear-10

169

Figure 5-19: Top floor displacements of primary structure coupled with DSFs with different bracket connector stiffness during 1940 El-Centro Earthquake

The above graph again shows the efficiency of the flexible connections in El

Centro earthquake. After approximately few seconds of the El Centro earthquake, the

structure with Shear-50 connections began to significantly reduce response of the

main structure. The reduction continued up to about 30 seconds.

Figure 5-20: Top floor displacements of primary structure coupled with DSFs with different bracket connector stiffness during 1995 Kobe Earthquake

Figure 5-20 distinctly displays that the incorporation of shear damping connections

to the structure façade has significantly changed the effects of the seismic loading on

the behaviour of the building system and produced more desirable results.

-200

-150

-100

-50

0

50

100

150

200

0 5 10 15 20 25 30

Lat

eral

dis

palc

emen

t (m

m)

Time (Sec)


-250-200-150-100-50

050

100150200250

0 5 10 15 20 25 30

Lat

eral

dis

palc

emen

t (m

m)

Time (Sec)


170

Figure 5-21: Top floor displacements of primary structure coupled with DSFs with different bracket connector stiffness during 1968 Hachinohe Earthquake

A similar trend was observed in the case of 1968 Hachinohe earthquake in Figure

5-21. The results showed that the top displacement of main structure is reduced under

this earthquake. From the results above, it appears that the consideration of movable

cladding reduces the top floor displacement of the main frame. According to these

results, it is seen that by selecting optimum value of shear stiffness brackets, the

overall lateral displacement of the primary structure subjected to seismic load is

decreased. Table 5-9 shows comparison between effects of different shear connections

on the top displacement reduction of the primary structure. Top displacement can be

reduced in the range of 30%-%50% in comparison to “Rigid” and “Axial” bracket

connections. In general, the results of the investigation of the proposed damping

system have demonstrated an ability to reduce the seismic response of buildings by

placement of the damping devices within the building facade system.

Table 5-9: Comparison between maximum top floor displacements of primary structure coupled with DSFs with different bracket connector stiffness during the four earthquakes

Earthquake Type of bracket system used in double skin façade system Rigid Axial Shear

Northridge 132 112 65 (Shear-10)El Centro 175 142 103 (Shear-50)

Kobe 226 208 119 (Shear-20)Hachinohe 103 91 70 (Shear-50)

-200

-150

-100

-50

0

50

100

150

200

0 5 10 15 20

Lat

eral

dis

palc

emen

t (m

m)

Time (Sec)


171

Top floor displacements of primary structure with various shear façade bracket

stiffness during the four Earthquakes are compared and shown in Figure 5-22 to

Figure 5-25 and Table 5-9.

Figure 5-22: Top floor displacements of primary structure coupled with DSFs with different shear connector during 1994 Northridge Earthquake

Bracket elements with shear stiffness of 10 N/mm has similar frequency to

dominant frequency of Northridge excitation and it can be seen from Figure 5-22 that

this connection can produce the highest reduction of top displacement of the main

structure among other shear connections. It is seen that values of 1 and 20 N/mm have

negligible effect on response reduction compared to the value of 10 N/mm. To be

more precise, it has been concluded that the optimum stiffness range is from 5 N/mm

to 15 N/mm for the selected earthquake record.

-200

-150

-100

-50

0

50

100

150

200

0 5 10 15 20 25 30

Lat

eral

dis

palc

emen

t (m

m)

Time (Sec)

Shear-100 Shear-50 Shear-10

-200

-150

-100

-50

0

50

100

150

200

0 5 10 15 20 25 30

Lat

eral

dis

palc

emen

t (m

m)

Time (Sec)


172

Figure 5-23: Top floor displacements of primary structure coupled with DSFs with different shear connector during 1940 El-Centro Earthquake

During 1940 El-Centro excitation, bracket elements with shear stiffness of 50

N/mm have similar frequency to dominant frequency of the applied record.

Additionally, it can be seen from Figure 5-23 that this connection can achieve the

highest reduction of top displacement of the main structure among other shear

connections. It is seen that values of 1, 10 and 20 N/mm have less effect on response

reduction compared to the value of 50 N/mm. To be more precise, it has been

concluded that the optimum stiffness range is values between 30 N/mm to 60 N/mm

for the selected earthquake record.

-200

-150

-100

-50

0

50

100

150

200

0 5 10 15 20 25 30

Lat

eral

dis

palc

emen

t (m

m)

Time (Sec)


-200

-150

-100

-50

0

50

100

150

200

0 5 10 15 20 25 30

Lat

eral

dis

palc

emen

t (m

m)

Time (Sec)


173

Figure 5-24: Top floor displacements of primary structure coupled with DSFs with different shear connector during 1995 Kobe Earthquake

Bracket elements with shear stiffness of 20 N/mm has similar frequency to

dominant frequency of Kobe excitation and it can be seen from Figure 5-24 that this

connection can be responsible for the highest reduction of top displacement of the

main structure among other shear connections. It is seen that values of 1 and 10 N/mm

have promising effect on response reduction of primary structure but less effective

than the optimum value of 20 N/mm. To be more precise, it has been concluded that

the optimum stiffness range is from 15 N/mm to 25 N/mm for the selected earthquake

record.

-250-200-150-100-50

050

100150200250

0 5 10 15 20 25 30

Lat

eral

dis

palc

emen

t (m

m)

Time (Sec)


-250-200-150-100-50

050

100150200250

0 5 10 15 20 25 30

Lat

eral

dis

palc

emen

t (m

m)

Time (Sec)


174

Figure 5-25: Top floor displacements of primary structure coupled with DSFs with different shear connector during 1968 Hachinohe Earthquake

Finally, it is illustrated in Figure 5-25 that bracket elements with shear stiffness of 50

N/mm are most effective in reduction of top displacement of main structure as their

frequency of movement is similar to dominant frequency of Kobe excitation. It is seen

that values of 1 and 10 N/mm have negligible effect on response reduction compared

to the value of 20 N/mm and 50 N/mm. To be more precise, it has been concluded that

the optimum stiffness range is from 15 N/mm to 60 N/mm for the selected earthquake

record. According to these results, it is shown that by selecting the optimum value of

shear stiffness for brackets, the overall lateral displacement of the primary structure

subjected to seismic load is decreased. Table 5-10 shows comparison between effects

of different shear connections on the top displacement reduction of the primary

structure. The maximum top displacement has reduction of around 50% in comparison

to “Rigid and Axial bracket connections.

-200

-150

-100

-50

0

50

100

150

200

0 5 10 15 20

Lat

eral

dis

palc

emen

t (m

m)

Time (Sec)


-200

-150

-100

-50

0

50

100

150

200

0 5 10 15 20

Lat

eral

dis

palc

emen

t (m

m)

Time (Sec)


175

Table 5-10: Comparison between maximum top floor displacements of primary structure with various shear façade bracket stiffness during the four earthquakes

Earthquake

Type of bracket system used in double skin façade system Shear bracket (N/mm)

100 50 20 10 1

Northridge 109 87 83 65 76

El Centro 147 103 132 132 128

Kobe 197 156 119 148 138

Hachinohe 89 70 73 83 81

In general, the results of the investigation of the proposed damping system have

demonstrated an ability to reduce the seismic response of buildings by placement of

the damping devices within the building facade system.

5.3.2.2.2 Root Mean Square (RMS) of Top Displacement (mm)


quantity. To have a better insight into the performance of the proposed damper

system, RMS of top floor displacement using different bracket stiffness is compared in

Table 5-11 for the 3D models correspondingly under different earthquake excitations.

It is illustrated in the table that by using shear connection with stiffness of 10N/mm

the RMS value is almost halved compared to a system with fixed conventional façade

panels during 1994 Northridge earthquakes. From the results of this analysis, it can be

concluded that the inclusion of façade with lateral damping connections in the

structure would inevitably result in great reduction in the response of main structure

during various applied earthquakes.

176

Table 5-11: Root mean square of top displacement using different value of shear stiffness for the 3D models during the excitations

Type of Structure

Northridge El-Centro Kobe Hachinohe

RMS %

Reduction RMS

% Reduction

RMS %

Reduction RMS

% Reduction

Conventional 51.20 - 69.2 - 66.1 - 49.9 - Axial 40.25 21 56.9 18 62.2 7 40.9 18

Shear-100 37.46 27 51.5 26 47.5 29 40.4 19 Shear-50 29.98 41 28.9 58 37.7 43 25.9 48 Shear-20 27.61 46 43.6 37 26.9 59 28.5 43 Shear-10 21.51 58 41.5 40 33.3 49 36.1 28 Shear-1 25.13 51 40.2 42 31.1 53 35. 29

5.3.2.2.3 Relative Displacement of Façade and Main Structure

Facade panel net movement is related to differential displacement between facade

and the main frame. In order to have a better understanding of the relative

displacement of the façade system, the 3D model results are shown in Table 5-12 to

Table 5-15. These levels of relative displacements are deemed practical and can be

accommodated in the design of double skin façade systems.

Table 5-12: Relative Displacement between the primary structure model and outer layer of façade system during 1994 Northridge earthquake

Storey Level

Relative Displacement (mm) Shear bracket (N/mm)

100 50 20 10 1

10 30.1 77.3 127.6 159.5 265.59 28.1 71.8 117.5 145.6 251.1 8 26.7 65.4 105.8 131.1 249.37 22.6 56.8 90.3 111.6 244.46 21.0 52.3 82.1 101.4 232.15 19.1 47.1 72.7 89.9 212.14 17.3 40.5 60.7 73.4 189.93 15.1 38.9 57.8 69.7 156.72 13.5 31.7 44.7 53.4 121.41 12.8 29.8 41.3 49.1 111.9

177

Table 5-13: Relative Displacement between the primary structure model and outer layer of façade system during 1940 El-Centro earthquake

Storey Level


100 50 20 10 1

10 39.1 100.5 165.9 207.4 345.29 37.5 94.3 153.7 190.4 337.38 36.1 86.3 138.9 171.7 325.4 7 28.4 72.9 116.4 144.2 316.76 27.4 68.8 106.8 131.9 301.85 25.8 62.3 95.7 117.6 276.74 22.5 52.7 79.03 95.5 246.93 19.0 49.9 74.4 90.0 202.92 18.1 41.8 58.7 70.1 158.41 16.7 38.8 53.6 63.8 145.4

Table 5-14: Relative Displacement between the primary structure model and outer layer of façade system during 1995 Kobe earthquake

Storey Level


100 50 20 10 1

10 48.9 125.6 207.3 259.2 431.59 44.9 115.9 190.2 235.9 407.18 39.1 101.8 167.6 208.6 400.77 31.6 87.1 141.5 176.2 391.96 28.2 79.1 127.5 158.9 371.35 25.3 70.9 112.4 140.0 338.84 23.2 60.9 93.7 114.3 303.63 19.7 58.4 89.0 108.5 249.62 17.7 47.2 68.4 82.5 193.01 16.1 44.5 63.1 75.8 177.9

Table 5-15: Relative Displacement between the primary structure model and out layer of façade system during 1968 Hachinohe earthquake

Storey Level

Relative Displacement Shear bracket (N/mm)

100 50 20 10 1

10 21.0 54.0 89.1 111.4 185.59 19.3 49.8 81.8 101.4 175.08 16.8 43.8 72.1 89.7 172.37 13.5 37.4 60.9 75.7 168.56 12.1 34.0 54.8 68.3 159.65 10.9 30.4 48.3 60.2 145.64 9.9 26.2 40.3 49.2 130.53 8.4 25.1 38.2 46.6 107.3 2 7.6 20.3 29.4 35.4 82.91 7.2 19.1 27.1 32.6 66.5

178

5.3.2.2.4 Structural Inter-Storey Drift

Another very important engineering demand parameter (EDP) for multistorey

concrete frame buildings is the evaluation of storey drifts. The interstorey drift ratio is

an important EDP because it helps to describe global damage to drift sensitive

components of the building such as structural framing, interior partitions, exterior

cladding, and window glazing. In this section, dynamic time-history analyses of the

model were performed to determine the maximum interstorey drift demand in each

storey.

5.3.2.2.4.1 Inter-Storey Drift Calculation and In-Plane Seismic Design

Buildings subjected to seismic excitations experience reverse cyclic swaying and the

resulting deformations induced in buildings may be quantified for the assessment of

façade systems using the inter-storey displacement (Calvi, Pinho et al. 2006). A

schematic diagram of a building sway under earthquake ground motion is illustrated in

Figure 5-26. The inter-storey drift ratio “∆I” at the “ith” floor can be defined as:

∆ ∆ )*x100% (5-1)

where “hi” is the storey height

According to clauses 5.4.4 and 5.5.4 of AS 1170.4 (2007), the inter-storey drift that

is calculated from the forces determined according to strength and stability provisions

at the ultimate limit state shall not be greater than 1.5% of the storey height for each

level (Mwafy and Elnashai 2001). Also cladding and façade panel’s attachment in a

seismic-force-resisting system shall have sufficient deformation and rotational

capacity. Therefore, for a typical floor height of 3600 mm, the maximum allowable

relative storey drift is 54 mm.

179

Figure 5-26: Schematic diagram of a building movement under earthquake ground motion.

5.3.2.2.4.2 Results

The interstorey drifts for the model were determined for each time-history analysis

and shown in Figure 5-27 to Figure 5-30. Additionally, absolute values of interstorey

drifts of each floor of the building with different brackets are shown in Table 5-16 to

Table 5-19 for all four selected records. Finally, absolute maximum value of

interstorey drifts are compared for each bracket case and shown in Table 5-20. The

interstorey drift ratios for all stories are plotted for the Northridge ground motions in

Figure 5-27. It can be seen that the maximum reduction of interstorey drifts is

achieved with connection of “Shear-10” in the 10th floor.

∆i

Hb

hi

180

Figure 5-27: Drift for primary structure with different stiffness for shear bracket façade elements during 1994 Northridge earthquake

From the results above, it appears that the consideration of movable cladding

reduces the maximum interstorey drift of the frame, especially for large intensity

earthquakes. However, to confirm these results, trends must be determined

considering all ground motions. The interstorey drift ratios in all stories are plotted

for the 1940 El-Centro ground motion in Figure 5-28. It can be seen that the maximum

reduction of interstorey drifts is achieved with connection of “Shear-50” in 10th floor.

Figure 5-28: Drift for primary structure with different stiffness for shear bracket façade elements during 1940 El-Centro earthquake

The results of the time-history show that the interstorey drifts in the moment-frame

are reduced significantly by the flexible shear connectors. These connections are the

most effective ones because of their stiffness and strength. They do deform or become

significantly damaged to absorb as much as the applied seismic energy. The

interstorey drift ratios in all stories are plotted for the 1995 Kobe ground motions in

0

10

20

30

40

50

1 2 3 4 5 6 7 8 9 10

Dri

ft b

etw

een

stor

eis(

mm

)

Storey

Conventional k=100 k=50 k=20 k=10 k=1

0

10

20

30

40

50

1 2 3 4 5 6 7 8 9 10

Dri

ft b

etw

een

stor

eis(

mm

)

Storey


181

Figure 5-29. It can be seen that the maximum reduction of interstorey drifts is

achieved with connection of “Shear-20” in 10th floor.

Figure 5-29: Drift for primary structure with different stiffness for shear bracket façade elements during 1995 Kobe earthquake

The frame model with bracket façades moving in direction of earthquake have

approximately the same results of fixed bracket, while frame model with bracket

façades moving in perpendicular to direction of earthquake has larger reduction in all

storey drifts. The interstorey drift ratios in all stories are plotted for the 1968

Hachinohe ground motions in Figure 5-30. It can be seen that the maximum reduction

of interstorey drifts is achieved with connection of “Shear-50” in 10th floor.

Figure 5-30: Drift for primary structure with different stiffness for shear bracket façade elements during 1968 Hachinohe earthquake

It is concluded from the above figures that, façade panels with energy absorbing

connections have a favourable effect on the overall structural behaviour and are able

to reduce interstorey drifts. Results of above figures are listed in Table 5-16 to Table

5-19 for better understanding of damper effects.

0

10

20

30

40

50

1 2 3 4 5 6 7 8 9 10

Dri

ft b

etw

een

stor

eis(

mm

)

Storey


0

3

6

9

12

15

1 2 3 4 5 6 7 8 9 10

Dri

ft b

etw

een

stor

eis(

mm

)

Storey


182

Table 5-16: Comparison of storey drift with different bracket stiffness during 1994 Northridge earthquake

Storey Level

Storey drift (mm) Shear bracket (N/mm)

Rigid 100 50 20 10 1

10 26.9 25.5 24.2 22.4 20.4 21.89 32.2 30.2 28.9 26.8 24.4 26.18 25.7 23.2 22.0 19.2 17.5 18.77 27.9 26.0 24.6 23.3 21.2 22.76 29.2 26.9 24.8 21.2 19.3 20.65 28.4 25.8 23.5 20.6 18.7 20.04 29.1 26.5 23.3 19.8 18.0 19.33 29.3 26.6 23.4 19.3 17.5 18.82 29.0 26.4 23.2 15.9 14.5 15.51 24.1 21.4 19.2 13.2 12.0 12.9

Table 5-17: Comparison of storey drift with different bracket stiffness during 1940 El Centro earthquake

Storey Level


Rigid 100 50 20 10 1

10 29.6 28.7 26.4 27.7 28.8 28.39 38.8 37.6 34.5 36.2 37.7 37.18 33.7 32.7 26.9 28.3 29.4 28.97 37.5 37.1 28.5 29.9 32.3 30.66 38.5 32.7 28.5 31.3 33.8 33.55 36.4 30.9 26.5 29.5 31.2 33.34 36.4 30.9 26.6 30.3 32.7 34.23 36.3 30.8 23.6 26.1 28.8 29.52 36.9 31.3 21.7 25.6 28.2 29.01 30.8 26.2 15.4 18.3 20.2 20.7

Table 5-18: Comparison of storey drift with different bracket stiffness during 1995 Kobe earthquake

Storey Level


Rigid 100 50 20 10 1

10 34.7 31.2 29.7 27.0 28.5 30.29 44.0 39.6 37.4 34.0 35.9 38.08 37.2 33.5 29.8 27.1 28.7 30.47 43.0 38.7 33.9 30.5 32.5 34.26 47.8 42.8 37.9 33.7 36.3 37.85 47.3 42.6 35.9 32.6 34.5 36.64 47.4 41.7 35.9 32.7 34.5 36.63 46.1 40.1 35.0 31.8 33.6 35.62 44.7 38.9 33.9 30.8 32.5 34.51 36.6 31.8 27.0 24.5 25.9 27.5

183

Table 5-19: Comparison of storey drift with different bracket stiffness during 1968 Hachinohe earthquake

Storey Level


Rigid 100 50 20 10 1

10 9.5 8.9 8.6 8.8 8.8 9.09 12.1 11.2 10.8 11.0 11.1 11.38 9.2 8.3 8.0 8.4 8.5 8.67 10.9 9.8 9.5 10.0 10.1 10.66 11.5 10.3 9.8 10.2 10.3 10.85 11.9 10.6 9.5 10.0 10.1 10.64 11.3 10.1 8.6 9.0 9.1 9.63 10.6 9.3 7.9 8.3 8.4 8.82 10.2 9.1 7.4 7.7 7.8 8.21 8.4 7.4 5.8 6.1 6.1 6.5

Absolute maximum value of interstorey drifts for each of the earthquakes are

compared for each bracket case in Table 5-20.

Table 5-20: Comparison of absolute maximum value of interstorey drifts for each bracket case

Earthquake


Rigid 100 50 20 10 1

Northridge 32.2 30.3 28.9 23.3 24.5 22.7

El Centro 38.8 37.7 34.5 36.2 37.7 37.1

Kobe 47.6 42.8 37. 9 33.7 36.4 38.1

Hachinohe 12.1 11.2 10.9 11.1 11.2 11.3s

5.3.2.2.5 Top Lateral Acceleration

The other global engineering demand parameter considered in this study is the

maximum floor acceleration. Floor accelerations are used to predict the damage to

acceleration sensitive components in the building, such as ceiling systems, chimneys,

and mechanical and electrical equipment. Various degrees of effectiveness of the

damping system with various stiffness of the connections for the various earthquake

records were studied. Response history of top floor acceleration of the three-

dimensional 10-storey building structure is presented in this section. Top floor

acceleration in case of rigid, axial and shear connections are extracted and compared

in various figures and tables of this section. Comparison of responses for the structure

184

with rigid bracket facade and structure with axial bracket facade showed that the

proposed connectors were not able to reduce the peak values of upper floor

acceleration. However, comparison of responses for the structure with rigid bracket

facade and structure with flexible shear bracket facade showed that the advanced

connectors were able to reduce peak values of upper floor acceleration. For each of the

records the connection with highest effects in reduction of top floor acceleration has

been selected in order to illustrate the performance of proposed system.

Figure 5-31: Time-history of top floor accelerations of primary structure coupled with DSFs with optimal bracket connector stiffness during 1994 Northridge Earthquake

From Figure 5-31 , it can be seen that the reduction in top acceleration with axial

connection is negligible. However, after approximately four seconds of the Northridge

earthquake, the structure with flexible shear façade system began to reduce the

response of the structure. The reduction is continued up to the end of the excitation.

The results for the same investigated parameter obtained by the frame structure with

same connections in façade brackets under 1940 El-Centro earthquake excitation are

presented in Figure 5-32.

-60

-45

-30

-15

0

15

30

45

60

0 5 10 15 20 25 30 35 40

Lat

eral

acc

eler

atio

n (m

m/s

2 )

Time (Sec)


185

Figure 5-32: Time-history top floor accelerations of primary structure coupled with DSFs with optimal bracket connector stiffness during 1940 El-Centro Earthquake

The above graph again shows the efficiency of the flexible connections. After

approximately three seconds of the El-Centro earthquake, the structure with Shear-50

connections began to reduce the response of the main structure. The reduction

continued up to end of the excitation.

Figure 5-33 : Time-history top floor accelerations of primary structure coupled with DSFs with optimal bracket connector stiffness during 1995 Kobe Earthquake

Figure 5-33 distinctly shows that the incorporation of shear damping connections to

the structure façade has changed the effect of the seismic loading on the behaviour of

the building system and produced desirable results.

-60

-45

-30

-15

0

15

30

45

60

0 5 10 15 20 25 30 35 40

Lat

eral

acc

eler

atio

n (m

m/s

2 )

Time (Sec)


-60

-45

-30

-15

0

15

30

45

60

0 5 10 15 20 25 30 35 40

Lat

eral

acc

eler

atio

n (m

m/s

2 )

Time (Sec)


186

Figure 5-34: Time-history top floor accelerations of primary structure coupled with DSFs with optimal bracket connector stiffness during Hachinohe Earthquake

A similar trend was observed in the case of 1968 Hachinohe earthquake shown in

Figure 5-34. The results showed that the top displacement of main structure is reduced

under this earthquake. From the results above, it appears that the consideration of

movable cladding reduces the top floor acceleration of the frame. Table 5-21 shows

the efficiency of the proposed damping connections in all stories.

Table 5-21: Comparison between top floor accelerations (mm/sec2) of primary structure coupled with DSFs with optimal bracket connector stiffness during 1994 Northridge Earthquake

Storey Level Type of bracket system used in double skin façade system

Rigid Axial Shear-1010 34.3 30.2 23.99 25.5 22.9 18.38 17.2 15.4 12.77 17.6 15.9 13.06 22.3 20.1 16.15 23.8 21.4 18.14 23.4 21.0 17.93 21.1 19.0 17.42 16.8 15.6 14.41 9.9 9.3 8.7

As can be seen from Table 5-22, the system achieved a very high level of

efficiency especially in upper stories.

-30-24-18-12-606

12182430

0 5 10 15 20

Lat

eral

acc

eler

atio

n (m

m/s

2 )

Time (Sec)


187

Table 5-22: Comparison between top floor accelerations (mm/sec2) of primary structure coupled with DSFs with optimal bracket connector stiffness during 1940 El Centro Earthquake



The above table shows high efficiency of the flexible connections in the upper

storey in terms of reduction of top lateral acceleration.

Table 5-23: Comparison between top floor accelerations (mm/sec2) of primary structure coupled with DSFs with optimal bracket connector stiffness during 1995 Kobe Earthquake



Table 5-23 and Table 5-24 show that top floor acceleration of main structure can be

reduced by using appropriate shear bracket in façade connections.

188

Table 5-24: Comparison between top floor accelerations (mm/sec2) of primary structure coupled with DSFs with optimal bracket connector stiffness during 1968 Hachinohe Earthquake



Top floor displacements of primary structure with various shear façade bracket

stiffness during the four Earthquakes are compared and shown in Table 5-25 to Table

5-28.

Table 5-25: Comparison between top floor accelerations (mm/sec2) of primary structurewith various shear façade bracket stiffness during 1994 Northridge earthquake

Storey Level


100 50 20 10 1

10 29.6 27.1 24.0 23.9 25.29 22.4 20.6 18.4 18.3 19.48 14.9 13.9 12.8 12.7 13.27 15.3 14.3 13.1 13.0 13.56 19.5 18.1 16.3 16.1 16.85 20.9 19.3 18.3 18.1 18.94 20.5 18.9 18.0 17.9 18.63 18.4 17.1 17.7 17.4 18.12 15.2 14.0 14.6 14.4 15.01 9.2 9.3 9.1 8.7 9.1

189

Table 5-26: Comparison between top floor accelerations (mm/sec2) of primary structure with various shear façade bracket stiffness during 1940 El Centro earthquake

Storey Level


100 50 20 10 110 36.3 29.4 30.6 31.2 33.99 31.1 26.8 27.9 28.4 30.88 23.4 20.7 21.6 22.0 23.97 20.9 20.2 21.1 21.5 23.36 21.1 21.6 22.5 22.9 24.85 20.2 20.1 20.9 21.3 23.24 16.4 16.2 16.9 17.2 18.73 14.3 14.2 14.8 15.1 16.42 11.5 12.0 12.5 12.7 12.61 9.3 9.1 9.1 9.2 9.2

Table 5-27: Comparison between top floor accelerations (mm/sec2) of primary structure with various shear façade bracket stiffness during 1995 Kobe earthquake

Storey Level


100 50 20 10 110 48.6 37.4 35.9 38.1 40.99 36.7 30.9 29.7 31.6 30.98 35.9 32.4 31.1 33.0 30.27 32.3 29.5 28.4 30.2 27.26 28.9 26.8 25.7 27.3 24.35 26.5 25.2 24.2 25.7 24.84 21.6 21.5 20.7 22.0 20.03 20.3 21.0 20.2 21.4 19.62 15.2 16.3 15.6 16.6 16.21 9.7 9.7 9.5 9.8 9.8

Table 5-28: Comparison between top floor accelerations (mm/sec2) of primary structure with various shear façade bracket stiffness during 1968 Hachinohe earthquake

Storey Level


100 50 20 10 110 22.1 17.2 18.4 19.1 20.89 20.6 17.0 18.2 18.9 19.48 20.8 17.5 18.7 19.4 19.57 21.1 16.7 17.9 18.6 19.86 21.7 17.2 18.4 19.1 20.45 22.8 18.1 19.3 20.1 21.44 24.6 19.6 20.8 21.6 23.23 26.7 24.3 25.7 26.7 25.12 26.8 24.9 26.3 26.8 25.21 19.9 18.9 19.6 19.7 19.6

It can be concluded that the flexible shear damping connections achieved excellent

reductions of top acceleration for all the earthquake excitations with the reductions

190

being slightly higher for the Kobe earthquake excitation than the other three. The

results showed that the lowest reduction in all investigated parameters was achieved

under the Hachinohe earthquake. In general, the results of the investigation of the

proposed damping system have demonstrated an ability to reduce the seismic response

of buildings by placement of the damping devices within the building facade system.

5.3.2.2.6 Root Mean Square (RMS) of Top Acceleration

To have a better appreciation of the damper performance, RMS of top floor

acceleration using different bracket stiffness is compared in Table 5-29 for the 3D

model for different earthquake excitations. It is seen in Table 5-29 that by using shear

stiffness of 50N/mm the RMS value is almost halved compared to a system with fixed

conventional façade panels during Hachinohe earthquake.

Table 5-29: Root mean square of top acceleration using optimal values of shear stiffness for the 3D models during the four excitations

Type of Structure


RMS %

Reduction RMS

% Reduction

RMS %

Reduction RMS

% Reduction

Conventional 11.1 - 12.3 - 16.0 - 5.7 - Axial 10.0 9.6 11.1 11.9 14.6 9.03 5.4 4.05

Shear-100 9.7 12.2 10.7 14.6 13.8 13.4 4.9 12.1 Shear-50 8.9 19.9 8.7 31.1 11.1 31.2 2.9 48.6 Shear-20 8.1 27.3 8.9 28.8 9.7 39.5 3.3 41.6 Shear-10 7.7 29.8 9.1 28.0 10.2 36.5 3.5 38.4 Shear-1 7.9 28.0 9.5 24.8 10.7 32.8 3.8 33.1

This reduction in value of RMS is clear in other three earthquake excitations as

well. Maximum reduction happens with “Shear –10”, “Shear –50” and “Shear –20” in

Northridge, El-Centro, and Kobe earthquakes, respectively. From the result of this

analysis, it can be concluded that the inclusion of façade with lateral damping

connections in the structure would inevitably result in great reduction in the response

of the main structure during various earthquakes.

191

5.3.2.2.7 Base Shear

Base shear is an estimate of the maximum anticipated lateral force that happens due

to seismic ground motion at the base of a structure. Base shear directly depends on the

input seismic acceleration and its value (V) is depends on below factors:

Soil conditions at the site

Proximity to potential sources of seismic activity (such as geological faults)

Probability of significant seismic ground motion

Level of ductility and over strength associated with various structuralconfigurations and the total weight of the structure

Fundamental (natural) period of vibration of the structure when subjected todynamic loading

Figure 5-35: Calculation of base shear in a structure

Base shear is an estimation of the total horizontal loads acting on the structure in a

"Static" time frame and can be calculated from below equation according to the

provisions of AS1170.4 (2007).

∑ (5.2)

Base shear forces of the 3D structural model are compared and tabulated in Table

5-30.

h1

W2

W1 F1

F3

h2 h3

F2

W3

192

Table 5-30: Comparison between base shear forces (Knaack, Klein et al.) of the 3-D primary structure in various bracket stiffness

Earthquake Type of bracket system used in double skin façade system

Rigid Shear-100 Shear-50 Shear-20 Shear-10 Shear-1

Northridge 1752 1644 1548 1040 990 1094

El Centro 2323 1920 1411 1599 1674 1870

Kobe 3001 2801 2450 1597 1611 1801

Hachinohe 1367 1219 929 1011 1078 1099

5.4 Findings And Conclusion

The use of energy absorbing connections (damping devices) in facade system to

mitigate the seismic applied force to a ten-storey building was investigated in this

chapter. The analytical results presented in Chapter 5 disclosed that the connection

properties had significant influence on dissipating seismic forces. A considerable

reduction in value of all parameters can be observed from the graphs and tables as the

movable dissipative façade brackets consistently reduced all investigated parameters

of the building structure by a reasonable margin. This chapter concludes that it is

feasible to design facade connections to mitigate part of applied force during seismic

events. The deformations within the shear connectors are drift-sensitive: the larger the

interstorey drift, the larger the deformations in the connections. After establishing the

feasibility of the procedure, the efficiency of the damping connection system and the

technique developed here is going to be investigated in a mid-rise 30 storeys structural

building model in order to have more general understanding of the system

performance.

193

CHAPTER 6

HIGH-RISE 30-STOREY

STRUCTRAL MODELS

194

6.1 Introduction

In this chapter, the performance of the proposed movable façade system models in

mid-rise building structure is comprehensively investigated using modal analyses and

nonlinear dynamic time-history analyses. The overall behaviour of the mid-rise

structural models with movable dissipative cladding is compared to the same structure

with conventional façade system, and the dynamic effects and behaviour of the new

cladding system are compared across the models with different axial and shear

stiffness connections. Analytical calculations in previous chapter have shown that the

movement of cladding system significantly influences the seismic response of mid -

rise structural buildings with movable dissipative cladding in moderate to severe

earthquakes.

In this chapter, the same procedure would be adopted to evaluate the proposed

moveable system in mid-rise building structures with a view to identify any major

differences in response of mid-rise vs high-rise structures with moveable facades and

subjected to earthquake loads. Effects of the movable cladding on the fundamental

period of the building are also discussed in this section. Additionally, effects of the

cladding on inter-storey drifts, floors displacement, floors acceleration, base shear and

plastic hinge rotations in beam/column connections are going to be discussed in this

chapter. Time-history analyses of each model are performed with selected recorded

ground motions to determine the structural reliability of the proposed connections.

Because, most of the modelling assumptions in this chapter are similar to Chapter 5, in

order to avoid repetition, description of some assumptions are not considered in this

chapter. These assumptions are as below:

Modelling approach and assumptions

195

Finite element analysis

Column, beam, slab, bracket and façade column mesh size

Direction of applied loads and building boundary conditions

Material properties

Consideration of seismic mass

Design of structural elements

Equivalent sections for structural beams and columns

Details of cladding system (double skin) modelling and their materialproperties

Façade column modelling

6.2 Assumptions

6.2.1 Element behaviour and structural modelling

In this section, behaviour of the proposed façade system is evaluated, employing a

three-dimensional (3-D) mid-rise structural model. Sectional dimensions used for the

model are listed in Table 6.1 and more details are given in Appendix-A.

Table 6-1: Structural sections for beam and column elements

Moment-resisting frame Storey/Floor Columns (mm) Storey/Floor Beams (mm)

24-30 400X400 26-30 300X40023 400X400

18-25 400X55021-22 450X450

20 500X50013-17 400X600

17-19 550X55012-16 600X600

9-12 400X60010-11 650X650

8-9 700X7001-8 400X600

1-7 800X800

Floor height and total building height are 3.6 meters and 129.6 meters, respectively.

196

Figure 6-1: Front view of exterior elevation of the 3D frame model

The front view of frame structure used for the 3D models in this section is shown in

Figure 6-1. The model has five bays in X and Z directions, respectively. The length of

each bay is considered to be 6 meters for both directions. It is assumed that building

mass is distributed equally at each floor along the height of the structure. Figure 6.2

shows beams and columns layout in the building structure plan.

Figure 6-2: Plan view of the 30m by 30m 3D frame model (dimensions are in mm)

6000 6000 6000 6000 6000

30000

6000

6000

6000

6000

6000

3000

0

Beam Plan

6000 6000 6000 6000 600030000

6000

6000

6000

6000

6000

3000

0

Colum n P lan

197

6.2.2 Details of Bracket modelling

In primary numerical modelling, the bracket elements were modelled using

COMBIN element, a spring-damper element, but, in order to have more accurate

results, these elements had been replaced by BEAM188 elements in detailed

numerical modelling. Material properties of the BEAM188 elements were defined in a

way that they represent and behave in the same way as a COMBIN element in bracket

modelling (axial and shear behaviour). Therefore, literally the bracket elements are a

beam element with spring behaviour. The “rigid” cladding system represents common

systems used in regular multistorey buildings in modern construction. The force-

deformation curve of the axial connection, which is for a hyper-elastic material, is

shown in Figure 6-3. Bracket elements in high-rise building structure are very similar

to mid-rise building structure.

Figure 6-3: Defined Force-deformation curve of the axial connection

General behaviour and concept that has been used for the bracket system in mid-

rise structural model has been developed in a similar fashion as in Chapter 5. The only

difference is the value of the second and third stiffness slopes. These values need to be

set, based on the value of the fundamental frequency of the primary structure, in order

F (

Kna

ack,

d (mm)

500-1000 N/mm

20 80 100

35-100 N/mm1-5 N/mm

K1

K2

K3

I II III

198

to affect the structure’s frequency and control the lateral displacement of façade

panels. Shear brackets consist of two connectors; one being the gravity connection

which behaves like a rigid link and attaches the outer skin of the façade system to the

structural slab and the other one is the damper connection which has a low stiffness in

axial direction. The force-deformation curve of the “shear” connection, which is the

modified version of axial connections, is shown in Figure 6-4. This connection is

modelled in a way that provides a rigid connection in compression but with much less

force resistance in lateral direction. The shear bracket consists of a single linear line

instead of multi linear stiffness.

Figure 6-4: Defined Force-deformation curve of the shear connection

6.3 Types of numerical modellings

6.3.1 Modal Analysis

Analysis of a building’s mode shapes gives an indication of the deformed shape of

the building during an earthquake. Depending on the frequency content of the ground

motion, one or more of the modes can be excited, and the deformed shape will then be

a superposition of the participating modes. A modal analysis was completed to

evaluate the effect of the cladding panels and their movable connections on the modal

periods and mode shapes of the 30-storey building. An analysis of a building’s

d (mm)

Linear Stiffness

F (

Kn)

K

199

vibration periods provides information on how the building might respond to lateral

excitation. Depending on the predominant frequency of the ground motion, a building

with a shorter (or longer) fundamental period might be subjected to higher (or lower)

intensity shaking.

6.3.1.1 Results

In this section, using more realistic modelling approaches, the fixed cladding

system does not significantly affect the vibration characteristics of the building. There

is only 5% difference between the fundamental period of the bare frame and system

with rigid bracket cladding. The first three vibration periods are given in Table 6-2 for

the models. The first mode shape of all models is almost linear, and the modal

ordinates are nearly identical.

Table 6-2: Modal vibration periods of models with bare-frame, fixed and flexible facades

Model Vibration Period (seconds)

Mode 1 Mode 2 Mode 3 Bare-frame 2.97 0.96 0.63

Fixed facade 2.82 0.91 0.62 Flexible facade 2.92 1.08 0.55

The effective modal mass percentage is a reliable measure on how much of the

total mass of the structure is participating in each vibration mode. The effective modal

mass percentages for the first three modes are provided in Table 6-3. The higher the

effective modal mass, the more dominant the lateral response of the building for an

arbitrary ground motion. However, there may be specific ground motions (near fault

motions), building sites (directivity effects, soft soil), or structural features (tall

buildings) where second or third mode are also participating just as much or possibly

more than the fundamental mode.

200

Table 6-3: Participating modal mass percentages

Model Participating modal mass percentages (%)

Mode 1 Mode 2 Mode 3 Bare-frame 81.7 11.34 5.75

Fixed facade 83.62 10.15 5.23 Flexible facade 75.93 16.38 5.68

The first three modes contribute to approximately 98-99% of the total lateral

response in all the analytical models in this study. This finding indicates that these

three modes capture essentially the entire lateral dynamic response of the building.

The first mode contributes over 80% to the total response, which is typical of well-

designed concrete frame buildings with an approximately uniform distribution of

structural properties and mass. Comparison of participating modal mass percentages

for the first three modes of the 3D structure is shown in Figure 6-5.

Figure 6-5: Comparison of effective modal mass percentages for the first three modes

6.3.2 Dynamic Time-History Analysis (Holistic Nonlinear time-history analysis)

The time-history analyses give better insight into how the building and cladding

system respond to the applied earthquake ground motions. The consequence of

nonlinear yielding, energy dissipation and damping of the cladding system are better

understood through time-history analyses. The seismic loadings applied to the

structural models in this study were from existing past earthquake records. These

81.7 83.675.9

11.3 10.216.4

5.8 5.2 5.7

0

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60

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Bare Frame Fixed Façade Frame Flexible façade

Par

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Mode 1

Mode 2

Mode 3

201

earthquake records are time histories of horizontal ground accelerations. Earthquakes

have various properties such as duration of strong motion, the range of dominant

frequencies and peak ground acceleration (PGA). For that reason, they would have

different effects on the structures. The range of dominant frequencies as well as the

duration of the strong motions were kept similar to achieve a consistent comparison of

the response of a structural model to different earthquakes. It should be mentioned that

dynamic time-history analyses were performed to further investigate the seismic

behaviour of both the structural frame and the cladding connectors. The analyses were

performed using the Newmark integration scheme and 2% Rayleigh damping was

assumed for all hazard levels. Dynamic interpretation of energy content of the selected

earthquakes (Power Spectral Density) is exactly the same as the one mentioned in

Chapter 5.

6.3.2.1 Judgement on engineering demand parameters

The global behaviour of the concrete moment resisting frame, due to earthquake

excitation, is typically described by maximum deformation, inter-storey drift ratios,

residual (permanent) drifts, floor accelerations, forces in bracket connections and base

shear force. These engineering demand parameters (EDPs) were determined during

the time history analyses of the analytical models and investigated in this section. In

this study, trends for maximum values of the engineering demand parameters, which

are typical of multi-storey concrete moment resisting frames, are discussed.

6.3.2.1.1 Top lateral displacement

Effects of a particular earthquake on a building structure are usually evaluated by

maximum values of displacement at the top level of the building structure. The

effectiveness of the damper system was studied with various connections stiffness

during various earthquake records. A comprehensive response history of top-level

displacement of the three-dimensional 30- storey building structure is presented in this

section. Top floor displacement, in the case of “Rigid”, “Axial” and “Shear”

connections, are extracted and compared in several figures and tables in this section.

Comparison of responses for the structure with rigid bracket facade and structure with

axial bracket facade showed that the proposed connectors were not able to reduce the

peak values of top floor displacement.

Figure 6-6: Time-history of top floor displacements of primary structure coupled with DSFs with different bracket connector stiffness during 1994 Northridge earthquake

Comparison of responses for the structure with rigid bracket facade and structure

with flexible shear bracket facade showed that the advanced connectors were able to

reduce peak values of top floor displacement. Shear-25 is chosen to compare with

rigid and axial bracket façade because it has the best performance and clearly

enhances the performance of the proposed system. From Figure 6-6, it can be

observed that the maximum values of displacement reduction occurred after 10

seconds of response. Between, 10 and 30 seconds, effectiveness of these kinds of

shear brackets are promising. In addition, compared to the shear brackets, the bracket

system with only axial façade system cannot reduce top lateral displacement of the

primary structure during applied earthquake excitation. Between 12 and 20 seconds,

some of top displacement peaks of the primary structure have negligible reduction.

The results for the same investigated parameter, obtained for the frame structure with

202

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the same connection properties of façade brackets under the 1940 El-Centro

earthquake excitation, are presented in Figure 6-7.

Figure 6-7: Time-history of top floor displacements of primary structure coupled with DSFs with different bracket connector stiffness during 1940 El Centro earthquake

The above graph again shows the efficiency of the flexible connections in El

Centro earthquake. After approximately few seconds of the El Centro earthquake, the

structure with Shear-80 connections began to significantly reduce the response of the

main structure. The reduction continued up to about 25 seconds. In the middle of this

earthquake excitation, response of the main structure reduces significantly with Shear-

80 connections.

Figure 6-8: Top floor displacements of primary structure coupled with DSFs with different bracket connector stiffness during 1995 Kobe Earthquake

Figure 6-8 distinctly displays that the incorporation of shear damping connections

to the structure façade has significantly changed the effects of the seismic loading on

the behaviour of the building system and produced more desirable results.

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Figure 6-9: Top floor displacements of primary structure coupled with DSFs with different bracket connector stiffness during 1968 Hachinohe Earthquake

A similar trend was observed in the case of 1968 Hachinohe earthquake in Figure

6-9. The results showed that the top displacement of main structure is reduced under

this earthquake. From the results above, it appears that the consideration of movable

cladding reduces the top floor displacement of the main frame. According to these

results, it is seen that by selecting the optimum value of shear stiffness brackets, the

overall lateral displacement of the primary structure subjected to seismic load is

decreased. Table 6-4 shows comparison between effects of different shear connections

on the top displacement reduction of the primary structure.

Table 6-4: Comparison between maximum top floor displacements of primary structure coupled with DSFs with different bracket connector stiffness during the four earthquakes

Earthquake Type of bracket system used in double skin façade system Rigid Axial Shear % of reduction

Northridge 544 490 281 (Shear-25) 48 El Centro 483 413 276 (Shear-80) 43

Kobe 691 620 419 (Shear-40) 39 Hachinohe 287 250 176 (Shear-80) 38

Top displacement can be reduced in the range of 35% - 50% in comparison to

“Rigid” and “Axial” bracket connections. In general, the results of the investigation of

the proposed damper system have demonstrated an ability to reduce the seismic

response of buildings by placement of damper devices within the building facade

system. Top floor displacements of primary structure with various shear façade

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bracket stiffness during the four Earthquakes are compared and shown in Figure 6-10

to Figure 6-13 and Table 6-5.

Figure 6-10: Top floor displacements of primary structure coupled with DSFs with different shear connectors during 1994 Northridge Earthquake

Bracket elements with shear stiffness of 25 N/mm displayed similar frequency to

dominant frequency of Northridge excitation and it can be seen from Figure 5-10 that

this connection can produce the highest reduction of top displacement of the main

structure among other shear connections. It is seen that values of 50 and 10 N/mm

have negligible effect on response reduction compared to the value of 25 N/mm. It is

concluded that the optimum stiffness range is from 10N/mm to 50 N/mm for the

selected earthquake record.

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Figure 6-11: Top floor displacements of primary structure coupled with DSFs with different shear connectors during 1940 El Centro Earthquake

During 1940 El-Centro excitation, bracket elements with shear stiffness of 80

N/mm have similar frequency to dominant frequency of the applied record.

Additionally, it can be seen from Figure 6-11 that this connection can achieve the

highest reduction of top displacement of the main structure among other shear

connectors. It is seen that values of 150 and 25 N/mm have less effect on response

reduction compared to the value of 100, 80 and 50 N/mm. It is concluded that the

optimum stiffness range is values between 50 N/mm to 100 N/mm for the selected

earthquake record. Bracket elements with shear stiffness of 40 N/mm has similar

frequency to dominant frequency of Kobe excitation and it can be seen from Figure 6-

12 that this connection can be responsible for the highest reduction of top

displacement of the main structure among other shear connections.

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Figure 6-12: Top floor displacements of primary structure coupled with DSFs with different shear connectors during 1995 Kobe Earthquake

Figure 6-13: Top floor displacements of primary structure coupled with DSFs with different shear connectors during 1968 Hachinohe Earthquake

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It is seen that value of 25 N/mm has promising effect on response reduction of

primary structure but is less effective than the optimum value of 40 N/mm. It is

concluded that the optimum stiffness range is from 20 N/mm to 50 N/mm for the

selected earthquake record. Finally, it is illustrated in Figure 5-13 that bracket

elements with shear stiffness of 80 N/mm are most effective in reduction of top

displacement of main structure as their frequency of movement is similar to dominant

frequency of Hachinohe excitation.

It is seen that values of 150, 40 and 25 N/mm have negligible effect on response

reduction compared to the value of 80 N/mm and 100 N/mm. It is therefore concluded

that the optimum stiffness range is from 60 N/mm to 90 N/mm for the selected

earthquake record. According to these results, it is shown that by selecting the

optimum value of shear stiffness for brackets, the overall lateral displacement of the

primary structure subjected to seismic load is decreased. Table 6-5 shows comparison

between effects of different shear connections on the top displacement reduction of

the primary structure. The maximum top displacement has reduction of around 50% in

comparison to “Rigid” and “Axial” bracket connections.

Table 6-5: Comparison between maximum top floor displacements (mm) of primary structure with various shear façade bracket stiffness during the four earthquakes

Earthquake

Type of bracket system used in double skin façade system Shear stiffness (N/mm)

150 100 80 50 40 25 10 1

Northridge 476 441 398 310 299 281 322 381

El Centro 380 298 276 282 320 324 333 367

Kobe 600 510 465 421 416 445 522 531

Hachinohe 247 211 175 224 239 250 255 264

In general, the results of the investigation of the proposed damper system have

demonstrated an ability to reduce the seismic response of buildings by placement of

damper devices within the building facade system.

209

6.3.2.1.2 Root mean square (RMS) response of top displacement (mm)


quantity. To have a better insight into the performance of the proposed damper

system, RMS response of top floor displacement using different bracket stiffness is

compared in Table 6-6 for the 3D model under different earthquake excitations. It is

illustrated in the table that by using shear connection with stiffness of 25 N/mm the

RMS value is almost halved compared to a system with fixed conventional façade

panels during 1994 Northridge earthquakes. The optimum values of shear bracket

stiffness are around 80,40,80 N/mm for El-Centro, Kobe and Hachinohe earthquakes,

respectively.

Table 6-6: Root mean square response of top displacement using different values of shear stiffness for the 3D models during different earthquake excitations

Type of Facade Bracket


RMS %

Reduction RMS

% Reduction

RMS %

Reduction RMS

% Reduction

Conventional 169 - 191 - 245 - 105 - Axial 150 11 171 10 224 8 97 8

Shear-150 146 14 151 21 220 10 90 14 Shear-100 134 21 97 49 183 25 74 30 Shear-80 109 35 89 53 137 47 63 40 Shear-50 73 56 99 48 121 51 78 26 Shear-40 70 58 104 46 116 52 81 23 Shear-25 68 59 125 34 129 47 94 10 Shear-10 76 55 129 32 187 23 94 10 Shear-1 102 39 149 22 190 22 96 9

From the results of this analysis, it can be concluded that the inclusion of façade

with lateral flexible connections in the structure would inevitably result in great

reductions in the response of the main structure during different earthquakes.

210

6.3.2.1.3 Maximum relative displacement of façade and main structure

Facade panel net movement is related to relative displacement between facade and

the main frame. In order to have a better understanding of the maximum relative

displacement of the façade system, the 3D model results are shown in Table 6-7 to

Table 6-10

Table 6-7: Maximum relative Displacement between 3D structure model and outer layer of façade system during 1994 Northridge earthquake

Storey Level

Maximum relative Displacement (mm) Shear Stiffness (N/mm)

100 50 25 10 1

30 77 197 397 413 63427 72 183 363 377 60524 68 167 326 340 58321 58 145 278 289 56318 54 133 253 263 53315 49 120 223 232 50612 42 98 174 187 4449 27 70 122 139 2706 21 49 80 92 1793 13 30 49 60 104

Table 6-8: Maximum relative Displacement between 3D structure model and outer layer of façade system during 1940 El-Centro earthquake

Storey Level


150 100 80 50 25

30 102 238 251 408 56927 98 224 236 379 52324 95 206 217 343 47321 77 175 184 289 39918 74 164 173 266 36615 70 151 159 239 32712 61 124 130 181 2689 39 87 91 152 2536 27 53 55 106 1993 20 39 41 84 182

211

Table 6-9: Maximum relative Displacement between 3D structure model and outer layer of façade system during 1995 Kobe earthquake

Storey Level


100 80 40 25 10

30 119 306 505 631 95527 109 282 463 575 90124 95 248 408 508 88721 77 212 345 429 86818 69 193 311 387 82215 62 173 274 341 75012 51 135 219 267 7109 34 101 153 187 4306 22 58 101 122 2853 17 44 62 75 175

Table 6-10: Maximum relative Displacement between 3D structure model and out layer of façade system during 1968 Hachinohe earthquake

Storey Level


150 100 80 50 25

30 50 128 143 233 29227 46 118 132 214 26524 40 104 116 189 23521 32 89 99 159 19818 29 81 90 144 17915 26 72 81 127 15812 23 66 63 97 1189 13 40 47 72 876 10 27 33 39 473 7 24 20 29 35

These levels of maximum relative displacements are deemed practical and can be

accommodated in the design of double skin façade systems.

6.3.2.1.4 Structural inter-storey drift

Another very important engineering demand parameter (EDP) for multistorey

concrete frame buildings is the evaluation of storey drifts. The inter-storey drift ratio

is an important EDP because it helps to describe global damage to drift sensitive

components of the building such as structural framing, interior partitions, exterior

cladding, and window glazing. In this section, dynamic time-history analyses of the

212

model were performed to determine the maximum inter-storey drift demand in each

storey. The maximum inter-storey drifts for the 3-D model were determined for each

time-history analysis and shown in Figure 6-14 to Figure 6-17. Additionally, absolute

maximum values of inter-storey drifts of each floor of the building with different

brackets are shown in Table 6-11 to Table 6-14 for all four selected earthquake

records. Finally, absolute maximum values of inter-storey drifts are compared for each

bracket case and shown in Table 6-15. The inter-storey drift ratios for all stories are

plotted for the Northridge ground motions in Figure 6-14. It can be seen that the

maximum reduction for inter-storey drifts is achieved with connection of “Shear-25”

in middle floors.

Figure 6-14: Maximum drift for primary structure with different stiffness for shear bracket façade elements during 1994 Northridge earthquake

From the results above, it appears that the consideration of movable cladding

reduces the maximum inter-storey drift of the frame, especially for large intensity

earthquakes. However, to confirm these results, trends must be determined

considering all ground motions. The inter-storey drift ratios in all stories are plotted

for the 1940 El-Centro ground motion in Figure 6-15. It can be seen that the maximum

reduction of inter-storey drifts is achieved with connection of “Shear-80” in 30th floor.

0

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Figure 6-15: Maximum drift for primary structure with different stiffness for shear bracket façade elements during 1940 El-Centro earthquake

The results of the time-history analyses show that the inter-storey drifts in the

moment resisting frame are reduced significantly by the flexible shear connectors.

These connections are the most effective ones because of their stiffness and strength.

They do deform or become significantly damaged to absorb as much as the applied

seismic energy. The inter-storey drift ratios in all stories are plotted for the 1995 Kobe

ground motion in Figure 6-16. It can be seen that the maximum reduction of inter-

storey drifts is achieved with connection of “Shear-40” in 30th floor.

Figure 6-16: Maximum drift for primary structure with different stiffness for shear bracket façade elements during 1995 Kobe earthquake

The frame model with bracket façades moving in direction of earthquake have

approximately the same results as fixed bracket, while frame model with bracket

façades moving perpendicular to direction of earthquake offer larger reduction for all

storey drifts. The Maximum inter-storey drift ratios in all stories are plotted for the

0

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1968 Hachinohe ground motion in Figure 6-17. It can be seen that the maximum

reduction of inter-storey drifts is achieved with connection of “Shear-80” in upper

floors.

Figure 6-17: Maximum drift for primary structure with different stiffness for shear bracket façade elements during 1968 Hachinohe earthquake

It is concluded from the above figures that, façade panels with energy absorbing

connections have a favourable effect on the overall structural behaviour and are able

to reduce inter-storey drifts. Results for the above figures are listed in Table 6-11 to

Table 6-14 for better understanding of damper effects.

Table 6-11: Comparison of maximum inter-storey drift with different bracket stiffness during 1994 Northridge earthquake

Storey Level

Maximum inter-storey drift (mm) Shear stiffness (N/mm)

100 50 25 10 1

30 27 26 23 26 2627 29 28 22 25 2624 23 22 18 20 2021 26 25 19 22 2318 26 25 20 22 2315 23 22 17 20 2012 23 22 17 20 209 25 25 19 22 226 22 21 17 19 203 20 20 15 17 18

0

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50

3 6 9 12 15 18 21 24 27 30

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Table 6-12: Comparison of maximum inter-storey drift with different bracket stiffness during 1940 El-Centro earthquake

Storey Level


150 100 80 50 25

30 24 23 24 24 2527 35 29 30 31 3224 34 32 31 34 3521 39 35 34 38 3918 47 38 37 40 4115 46 33 31 36 3712 47 32 31 36 379 44 32 30 36 376 42 30 28 35 363 34 22 21 26 27

Table 6-13: Comparison of maximum inter-storey drift with different bracket stiffness during 1995 Kobe earthquake

Storey Level


100 80 40 25 10

30 34 34 32 33 3427 42 40 38 39 4124 43 44 42 43 4521 53 53 51 52 5418 55 53 51 55 5415 54 45 42 45 4712 53 47 43 46 489 50 47 43 46 496 50 46 39 42 453 41 39 31 33 37

Table 6-14: Comparison of maximum inter storey drift with different bracket stiffness during 1968 Hachinohe earthquake

Storey Level

Maximum inter-storey drift (mm) Shear Stiffness (N/mm)

150 100 80 50 25

30 17 16 17 16 1627 18 18 18 18 1824 18 18 18 18 1821 22 22 22 22 2218 23 21 22 23 2315 23 20 21 22 2212 23 19 20 22 229 22 18 19 21 226 21 17 17 20 203 17 14 14 16 17

Absolute maximum values of inter-storey drifts for each of the earthquakes are

compared for each bracket case in Table 6-15.

216

Table 6-15: Comparison of absolute maximum values of inter-storey drifts for each bracket case

Earthquake

Type of bracket system used in double skin façade system Shear Stiffness (N/mm)

Rigid 150 100 80 50 40 25 10 1

Northridge 37 31 29 29 28 23 20 26 22

El Centro 54 47 38 37 40 40 41 43 43

Kobe 64 60 55 53 53 51 55 54 55

Hachinohe 26 23 22 22 23 23 23 23 23

6.3.2.1.5 Top lateral acceleration

The other global engineering demand parameter considered in this study is the

maximum floor acceleration. Floor accelerations are used to predict the damage to

acceleration sensitive components in the building, such as ceiling systems, chimneys,

and mechanical and electrical equipment. Various degrees of effectiveness of the

damper system with various stiffness of the connections for the various earthquake

records were studied. Response history of top floor acceleration of the three-

dimensional 30- storey building structure is presented in this section. Top floor

acceleration in case of rigid, axial and shear connections are extracted and compared

in various figures and tables of this section. Comparison of responses for the structure

with rigid bracket facade and structure with axial bracket facade showed that the

proposed connectors were not able to reduce the peak values of upper floor

acceleration. However, comparison of responses for the structure with rigid bracket

facade and structure with flexible shear bracket facade showed that the advanced

connectors were able to reduce peak values of upper floor acceleration. For each of the

records the connection with highest effects in reduction of top floor acceleration has

been selected in order to illustrate the performance of the proposed system.

217

Figure 6-18: Time-history of top floor accelerations of primary structure coupled with DSFs with optimal bracket connector stiffness during 1994 Northridge Earthquake

From Figure 6-18, it can be seen that the reduction in top acceleration with axial

connection is negligible. However, after approximately 4 seconds of the Northridge

earthquake, the structure with flexible shear façade system began to reduce the

response of the structure. The reduction is continued up to the end of the excitation.

The results for the same investigated parameter obtained by the frame structure with

the same connections in façade brackets under 1940 El-Centro earthquake excitation

are presented in Figure 6-19.

Figure 6-19: Time-history top floor accelerations of primary structure coupled with DSFs with optimal bracket connector stiffness during 1940 El-Centro Earthquake

The above graph again shows the efficiency of the flexible connections. After

approximately three seconds of the El Centro earthquake, the structure with Shear-80

connections began to reduce the response of the main structure. The reduction

continued up to end of the excitation.

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Figure 6-20: Time-history top floor accelerations of primary structure coupled with DSFs with optimal bracket connector stiffness during 1995 Kobe Earthquake

Figure 6-20 distinctly shows that the incorporation of shear damper connections to

the structure façade has changed the effect of the seismic loading on the behaviour of

the building system and produced desirable results.

Figure 6-21: Time-history top floor accelerations of primary structure coupled with DSFs with optimal bracket connector stiffness during Hachinohe Earthquake

A similar trend was observed in the case of 1968 Hachinohe earthquake shown in

Figure 6-21. The results showed that the top displacement of main structure is reduced

under this earthquake. From the results above, it appears that the consideration of

movable cladding reduces the top floor acceleration of the frame. Table 6-16 to Table

6-19 show the efficiency of the proposed damper connections in every other three

stories of the primary structure for the four ground motions.

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Table 6-16: Comparison between top floor accelerations (mm/sec2) of primary structure coupled with DSFs with optimal bracket connector stiffness during 1994 Northridge Earthquake


Rigid Axial Shear-1030 86 81 5827 63 60 4224 51 49 3321 42 39 2818 45 43 2915 41 39 3012 44 41 309 47 44 336 32 30 233 18 17 15

As can be seen from Table 6-16, the system achieved a very high level of

efficiency especially in upper stories.

Table 6-17: Comparison between top floor accelerations (mm/sec2) of primary structure coupled with DSFs with optimal bracket connector stiffness during 1940 El Centro Earthquake


Rigid Axial Shear-1030 124 120 7927 112 109 6824 104 101 6321 96 93 5818 82 80 4915 72 70 4312 60 58 369 49 47 296 36 35 213 25 24 15

The above table shows high efficiency of the flexible connections in the upper

storey in terms of reduction of top lateral acceleration.

220

Table 6-18: Comparison between top floor accelerations (mm/sec2) of primary structure coupled with DSFs with optimal bracket connector stiffness during 1995 Kobe Earthquake


Rigid Axial Shear-10 30 156 136 108 27 139 120 104 24 130 113 88 21 112 97 82 18 105 91 78 15 101 88 70 12 94 82 61 9 68 59 52 6 50 43 40 3 36 31 27

Table 6-18 and Table 6-19 also show that top floor acceleration of main structure

can be reduced by using appropriate shear bracket in façade connections.

s.Table 6-19: Comparison between top floor accelerations (mm/sec2) of primary structure .Table 6-19:

Comparison between top floor accelerations (mm/sec ) of primary structure cou

coupled with Comparison between top floor accelerations (mm/sec ) of primary structure couDSFs with optimal bracket connector stiffness during 1968 Hachinohe Earthquake


Rigid Axial Shear-10 30 62 49 35 27 53 41 30 24 44 34 23 21 40 31 20 18 40 31 19 15 37 28 16 12 35 27 14 9 32 24 11 6 25 19 10 3 11 8 7

Top floor acceleration of primary structure with various shear façade bracket

stiffness during the four Earthquakes are compared and shown in Table 6-20 to Table

6-23.

221

Table 6-20: Comparison between top floor accelerations of primary structure with various shear façade bracket stiffness during 1994 Northridge earthquake

Storey Level

Top Acceleration (mm/sec2) Shear Stiffness (N/mm)

100 50 25 10 1

30 76 62 58 60 6427 57 45 42 45 4724 47 35 33 37 3721 37 30 28 29 3118 40 31 29 32 3215 37 32 30 29 3312 39 32 30 31 339 42 35 33 33 366 29 25 23 23 263 17 15 15 14 16

Table 6-21: Comparison between top floor accelerations of primary structure with various shear façade bracket stiffness during 1940 El Centro earthquake

Storey Level


150 100 80 50 25

30 115 83 79 83 8827 105 71 68 71 7524 97 66 63 66 7021 89 60 58 60 6418 76 52 49 52 5515 66 45 43 45 4812 54 38 36 38 409 44 31 29 31 326 31 23 21 23 243 23 16 15 16 17

Table 6-22: Comparison between top floor accelerations of primary structure with various shear façade bracket stiffness during 1995 Kobe earthquake

Storey Level


100 80 40 25 10

30 131 122 108 110 11927 123 116 104 106 11524 109 100 88 90 9821 94 93 82 84 9118 88 91 78 79 8615 85 82 70 71 7812 79 71 61 62 679 59 57 52 53 576 44 42 40 41 443 31 29 27 27 29

222

Table 6-23: Comparison between top floor accelerations of primary structure with various shear façade bracket stiffness during 1968 Hachinohe earthquake

Storey Level


150 100 80 50 25

30 48 41 35 35 3827 41 34 30 30 3224 34 29 23 31 3321 31 26 20 31 3318 31 26 19 32 3415 29 23 16 20 2312 27 16 14 18 219 25 14 11 15 196 20 12 10 14 183 8 7 7 11 13

It can be concluded that the flexible shear damper connections achieved excellent

reductions of top acceleration for all earthquake excitations with the reductions being

slightly higher for the Kobe earthquake excitation than the other three. The results

showed that the lowest reduction in all investigated parameters was achieved under

the Hachinohe earthquake. In general, the results of the investigation of the proposed

damper system have demonstrated the ability to reduce the seismic response of

buildings by placement of the damper devices within the building facade system.

6.3.2.1.6 Root mean square (RMS) of top acceleration response

To have a better appreciation of the damper performance, RMS of top floor

acceleration using different bracket stiffness is also compared in Table 6-24 for the 3D

model for different earthquake excitations. It is seen in Table 6-24 that by using shear

stiffness of 40N/mm the RMS value is almost halved compared to a system with fixed

conventional façade panels during Kobe earthquake.

223

Table 6-24: Root mean square of top acceleration (mm/sec2) using optimal values of shear stiffness for the 3D models during the four excitations

Type of Structure


RMS %

Reduction RMS

% Reduction

RMS %

Reduction RMS

% Reduction

Conventional 24.3 - 37.7 - 46.5 - 13.5 - Axial 22.7 7 35.3 6 43.0 8 12.9 4

Shear-150 22.0 9 35.0 7 41.0 12 10.5 22 Shear-100 21.4 11 25.1 33 39.2 16 8.80 34 Shear-80 18.5 24 18.0 52 36.4 22 7.10 47 Shear-50 17.4 28 25.2 33 24.2 48 8.10 40 Shear-40 13.6 44 25.5 32 20.7 55 8.20 39 Shear-25 12.6 48 26.7 29 32.7 30 8.50 37 Shear-10 16.9 30 29.6 21 35.6 24 8.80 35 Shear-1 18.1 26 30.2 20 38 18 9.10 32

This reduction in value of RMS is also seen in other three earthquake excitations as

well. Maximum reduction happens with “Shear –25”, “Shear –80” and “Shear –80”

for Northridge, El-Centro, and Hachinohe earthquakes, respectively. From the result

of this analysis, it can be concluded that the inclusion of façade with lateral damper

connections in the structure would inevitably result in great reduction in the response

of the main structure during various earthquakes.

6.3.2.1.7 Base shear force

Base shear force is an estimate of the maximum anticipated lateral force that

happens due to seismic ground motion at the base of a structure. Base shear force

values depend on some parameters as described, comprehensively, in Chapter 5.

Values of base shear. The values are compared in Table 6-25.

Table 6-25: Comparison between base shear (Knaack, Klein et al.) of the primary structure with various bracket shear stiffness

Earthquake

Type of bracket system used in double skin façade system Shear (N/mm)

Rigid 150 100 80 50 40 25 10 1

Northridge 9.5 9.1 8.1 7.5 6.9 6.7 6.2 6.4 6.6

El Centro 14.8 13.2 10.9 8.9 9.6 9.9 10.3 11.0 11.3

Kobe 17.8 16.4 15.5 13.4 12.5 12.1 12.3 12.8 13.3

Hachinohe 7.9 7.7 6.6 5.9 6.3 6.7 7.0 7.1 7.6

6.4 Application of Advanced Cladding Connections and Design Steps

The idea of advanced cladding connections developed in this research was created to

take advantage of energy dissipation due to the relative movement of the cladding panels

and structural frame. These systems necessitate cladding systems to encounter significant

lateral movements to create any promising effects; therefore, crucial criteria such as the

appearance, water tightness, and air tightness due to the relative panel-structure movement

could be major issues to overcome. On the other hand, From the several time-history

analyses carried out, it was evident that with the implementation of appropriate

connection properties, the dynamic response of the main structure can be considerably

reduced. Moreover, the connection deformation and the connection forces can be kept

within reasonable and practical limits by applying pre-defined load deformation behaviour

to them. Engineers, often prefer to optimize a system for a particular property such as low

energy consumption. Particularly in the design of buildings, it is difficult because of

conflicting priorities such as optimizing daylight and minimizing solar gain. Thus, co-

optimizing is essential. The relationship between architects, engineers, and facility

managers must be managed carefully to develop and use complicated control features that

do not overwhelm users or the lack of management may render the advanced facades

inefficient. Additional design tools must be developed so that structural and façade

designers may easily investigate effects of movable double-skin facades on seismic

performance of the structure during design stage. It is noteworthy to mention that,

224

225

acceptance of the proposed façade system is linked to the additional expense to the

building owner, also architectural and environmental benefits indicated to them.

6.5 Findings and conclusion

The use of energy absorbing connections (damper devices) in facade system to

mitigate the seismic applied force to a thirty-storey building was investigated in this

chapter. The analytical results presented in Chapter 6 indicated that the connection

properties have significant influence on dissipating seismic energy. It is essential to

have the right combination of stiffness and panel mass. Sensitivity analyses were carried

out on changes to stiffness value using low medium, and high stiffness. Up to nine

different stiffness values were chosen to cover most probable and achievable range of

stiffnesses. The supporting analysis, mainly in Chapter 4, identified a practical challenge,

namely requiring the panels to move several meters to be effective. In Chapters 5 and 6,

multi linear behaviour damper was used instead to control the large displacement of the

panels and also display a similar reduction in the response of the main structure. The

second slope (soft stiffness) of the force-displacement behaviour of the damper system

part in the façade damper behaviour system leads the system to behave like a multi tuned

mass damper system. A considerable reduction in value of all parameters can be

observed from the graphs and tables as the movable dissipative façade brackets

consistently reduced all investigated parameters of the building structure by a

reasonable margin. This chapter concludes that it is feasible to design facade

connections to mitigate part of applied energy during seismic events. The

deformations within the shear connectors are drift-sensitive: the larger the inter-storey

drift, the larger the deformations in the connections. By controlling the connector

stiffness and introducing variable stiffness, one could reduce the primary structure

response and also limit the movement of the outer skin of the façade to a practical value.

226

A damper system designed based on applied earthquake force, may have some limitations

in terms of beneficial effects. This system works in a range of frequency content and it

can be demonstrated, based on sensitivity analysis ,performed through different intensity

of earthquake load. Low stiffness path should be designed in a way that façade panels

reach the right frequency to dissipate the input energy. After establishing the feasibility

of the procedure, the efficiency of the damper connection system and the technique

developed here were evaluated by applying them to a mid-rise 30 storeys structural

building model in order to gain a better insight of the system performance. Stiffness

value of shear bracket facades should be selected in a range between 20 N/mm up to

90 N/mm in order to reduce response of primary structure in most cases of applied

excitations.

227

CHAPTER 7

FINANCIAL

CONSIDERATIONS

228

7.1 Introduction

One of the new methods that have been proposed for strengthening and reducing

seismic damage for insufficient earthquake resistance of existing building structures

would be retrofitting. Seismic retrofitting or rehabilitation is the modification or

making changes to the existing structures to make them more resistant to seismic

activity and increase safety for the building’s occupants.

Table 7-1: Existing methods of retrofitting

Available retrofit method

Usage Image

Conventional strengthening

Add walls, brace, columns or enlarging excising structural elements

Innovative material

High performance concrete High performance composites FRP composites

Base isolation Placing flexible isolation system

between foundation and structure

Energy dissipation system

Passive or active structural control to reduce vibration and improve

safety/serviceability

229

The basic concept of retrofitting is to upgrade the lateral strength and increase the

ductility of the structure. The earthquake damaged, and earthquake vulnerable

buildings are classified to be retrofitted by innovative methods tabulated in Table 7-1

7.2 Financial Assessment of movable façade system as a new retrofitting

method

Engineering demand parameters (EDPs) such as the inter-storey drift and the

maximum relative deformations of the cladding connectors are useful for engineers

and researchers. However, the end-users and decision-makers of the systems are not

able to translate these technical EDPs data into useful quantities for making decisions

on practicability and efficiency of the system. The financial feasibility of the

proposed system is going to be discussed in two ways. Firstly, the additional cost of

replacing the conventional bracket system with the proposed system will be discussed,

then advantages of material saving in the building structure will be evaluated and at

the end of this section, conclusions and basis for decision will be shown by comparing

these two costs.

7.3 Additional cost of the movable facade to building structure

7.3.1 Introduction

By introducing this state-of-the-art system, two additional expenses, namely design

and maintenance of the proposed system are expected to add to overall cost of the

building structure. As the system would be new to the façade industry and not many

engineers and technicians are familiar with the system, then, initial training and

workshops need to be conducted to train them for dealing with the new system. It

should be mentioned that, along with aesthetics, a façade must be designed to separate

the interior of a building from the aggressive exterior environment and it must

230

withstand the imposed mechanical and environmental loads. The service life of the

proposed system starts when the construction work is finished. At this stage, bracket

of facades, especially the ones that are designed for double skin façade systems, may

deteriorate during their service life due to effect of various aggressive environment

mechanisms along with applied repetitive environmental loads. Therefore, the

additional costs of replacing the conventional façade system with the proposed system

is going to be discussed in sections 7.3.2 and 7.3.3.

7.3.2 Design or re-design procedure

Depending on building age and request by the building owner, two possible

scenarios are expected to happen as below:

In the first scenario, dynamic analysis of an existing structure shows that the

structure is vulnerable to future earthquakes with a specific return period, and the

owner is concerned. Therefore, besides other possible retrofitting possibilities,

structure/façade contractor of the movable façade system would submit a quote and

the expected percent of dynamic behaviour improvement to the building owner. The

proposal and final submitted quote would be based on the amount of construction

work and replacement of the current façade system. Additionally, similar proposals

can be offered for older existing buildings, which use traditional stick façade system.

In structures with traditional brick veneer façade system with opening windows, the

“In-plane” façade movement concept proposed in Chapter 4 of the thesis can be used

as one of the proposals. But this idea needs more research and validation before being

proposed to industry and become commercially viable. Details of the proposal will be

presented as future work in the next chapter. The second scenario is for new buildings

and a quotation of costs would be submitted to building developer and structural

231

design team. The quote will include proposals about possible façade systems that are

capable of integrating the new bracket system and the amount of improvement and

financial savings in relation to the dynamic behaviour of the future structure and

structural materials (volume of pumped concrete and weight of steel), respectively.

The cost of the smart façade system is definitely higher than the traditional facade

system as the cost of the damper device and details of design or re-design would be

added to the cost of the normal façade system. Table 7-2 shows the difference between

prices of these two façade systems.

Table 7-2: Details of additional price of smart façade system

Item Assumed price

(Amadio and Bedon) Price per square meter of façade system

(AUD/m2) Smart damper

device 700 (per each

bracket 115

Details for re-design

240 (per façade unit) 40

7.3.3 Maintenance

7.3.3.1 Preventive maintenance strategies and their cost

Several progressive degradation mechanisms begin to have a negative effect on the

constructed new bracket materials. Therefore, to extend the service life of them,

planned maintenance must be conducted. The maintenance strategies are mainly

divided into preventive and corrective maintenance. Consideration of a regulated

planned maintenance strategy is necessary at this stage. It should be mentioned that if

the efficiency of the system is not considered and predicted at the design stage, then

maintainability issues will arise very soon. These external actions produce continuous

deterioration in each one of the façade elements, especially the ones located in the

outer skin. The actions can only be avoided through a durable design method for the

new system. Some maintenance strategies are proposed as below:

232

a) Various maintenance procedures need to be considered for implementation of the

façade bracket dampers during its service life, including the necessary preventive

and corrective maintenance.

b) The design process of the proposed system should be conducted in a practical and

efficient manner for the adapted maintenance strategy.

c) Maintenance techniques must be performed to guarantee the accomplishment of

the design service life for every component of the facade system.

d) Maintenance operations should be facilitated by the adoption of a simple

geometry for easy access for inspection of hidden parts of the damper components

such as connecting steel joints.

e) Licenced architects or engineers with sound knowledge of the design, material

and construction of the bracket façade should conduct inspections. Periodical

visual and detailed inspections must also be carried out on façade brackets to

ensure the integrity and safety of the proposed systems and if inefficiencies are

detected, corrective maintenance plans, according to the type of deficiency, must

be chosen.

f) Provide access for installation of any needed instruments for testing and

developing information about the future behaviour and performance of the

constructed materials and used components.

g) The building owner should check with the design team to reach an agreement on

selecting the best façade cleaning method that is suitable for their building.

Regular cleaning of the system for proper operation to avoid gradual build up of

deterioration sources (microclimates), especially during the summer season when

severe conditions like high temperature and humidity are dominant. Managing

these cleaning procedures to be performed from time to time is essential, as well

233

as providing guidelines for seasonal cleaning activities for critical zones that can

accumulate salts and other aggressive agents. The cleaning activities for the

damper elements must be considered in preventive maintenance procedures to

decrease the aggressiveness of the microclimates and these steps are as below:

• Cleaning of the expansion joints at the supports to clear the accumulated debris,

dust and water-borne aggressive agents.

• Eliminating or cleaning the accumulated debris, dust and water carrying

aggressive materials underneath the expansion joints.

• Cleaning the fixed and mobile bearings and to maintain them for suitable

conditions to ensure their efficient and proper performance.

7.3.3.2 Quarterly and annual inspection of the proposed system

A team of professionals need to inspect the damper system at least every three

months to check overall performance of the system. Movement of materials, thermal

movements, moisture movements, elastic deformations, creep effects and corrosion

are the issues to be addressed. All deficiencies noted in quarterly reports should be

documented and also included in a yearly report. In this way, primary cause and level

of severity of the issues can clearly be recognised. The severity of the identified

deficiencies can be classified into the following conditions:

1. “Unsafe” is used when the identified deficiency causes a serious threat to

individuals or property and should be immediately brought to the attention of owner(s)

and local authorities by providing potential repair and corrective options.

2. “Requires repair/stabilization” is used for a case that may become unsafe if it is

not scheduled for the next maintenance program.

3. “Ordinary maintenance” recognizes the cases when something is required to be

addressed for the next scheduled inspection program.

234

A proposed spreadsheet, which states condition of each damper component, is

shown in Table 7-3. This spreadsheet would assist the owner/strata manager with

budgeting of future maintenance of the façade connectors based on their severity

classification. The budget evaluation should include all costs of contractor’s labour,

materials, equipment, overhead, and general conditions, as well as the fees for

architecture and engineering services, building administration, and unpredicted events.

Spreadsheet for forecasted costs is listed in Table 7-4.

Table 7-3: Proposed quarterly and yearly spreadsheet for inspection of each damper/connector component

System components Severity classification (Inspected condition)

Unsafe Requires

repair/stabilization Ordinary

maintenance

Rubber Steel layer

Washer Visco-Elastic material

Façade to bracket attachment Bracket to slab attachment

To guide the decision makers for planning the necessary repairs and future

inspections, the report should include a comprehensive survey of history and condition

of the façade panels in a way that can be understood by non-technical people as well.

Table 7-4: Spreadsheet for expected yearly expenses per square meter of façade panel

Item Estimated cost (Amadio and

Bedon)

Contractor’s labour 600

Materials 1,100

Equipment 700

Overhead 300

Fees for architecture and engineering 1,000

Building administration 500

Unpredicted events 20% of all expenses

Total cost 3,840

235

Additionally, the original building construction, alterations, renovations, and

repairs should be included in making a more precise decision. The value for yearly

expenses of the smart façade is going to be compared with conventional façade system

within this chapter.

7.3.4 Importance of thermal performance

One of the main criteria to evaluate the overall performance of a façade system as

an outer skin of building structure is thermal performance. Maintaining the same

thermal performance similar to conventional façade system is an essential element to

convince the developers, building owns and insurance companies of its merits.

Configuration and placement proposal of the damper system in the cavity between

outer and inner layer is a crucial decision to minimize renovation expenses and quoted

construction price for existing and new building structures, respectively.

7.3.5 Damage cost to the façade system after an earthquake

If after an earthquake, the damage is minor, and some joints just need to be sealed,

then the best action would be caulking between wedge-shaped of the system. In this

way, the only issue is the thermal performance of the system, and it is still

dynamically ready for the next earthquake. But, widespread damage will be sustained

by the façade panels, damper system and bracket connection, by excessive movement

of the facade system during large earthquakes. Although saving peoples’ lives by

reducing damage in structural elements is the first object of this system but trying to

find a way to minimize damage in the façade panels is equally worthwhile. Because,

as mentioned before, normal cladding systems cost around 20% of building

construction cost and for movable façade systems, this cost is increased to 25-30% of

building construction cost. Calculated movement of a façade panel from one of the

236

previous dynamic time history analyses performed in Chapter 5 is compared with

maximum allowable movement in façade panel in Table 7-5. The allowable façade

panel movement is based on national and international codes of AS1288, AS2047,

AS4284 and BS8118. The kind of repair method that is going to be used after a major

earthquake is a key factor for rehabilitation of building fenestration.

Table 7-5: Damage cost and damage state of the movable façade system located on top level of a mid-rise structure

Components Façade movement (mm)

Level of damage to the system And estimated repair/replacement cost

Calculated Allowable Damage

prevention Life

safety Immediate

occupancy

Glazed Exterior 35 h/60 $900 Aluminium frame 20 h/250 $1200

Gasket 3 5 $100 Rebate - - $100Glazing

bead(sealant bead) - - $1400

Setting block - - $350 Cover cap - - $300

Glass spacer 3 5 $300

Total cost and overall damage state

$4650

It is noteworthy to state that repair or replacement cost for each component of the

façade system is calculated after discussions with façade experts and industry

professionals.

7.3.6 Interaction with insurance companies for Earthquake Insurance Premium

(EIP)

It is very important that a comprehensive report is prepared for submission to

insurance companies and relevant authorities. Building insurance premiums depend on

many factors, and the most important ones are listed below:

Probability prediction of seismic hazard at the building location

Seismic design of the building

237

Building damage

Figure 7-1 shows changes of earthquake insurance premium versus seismic zone

where the structure is located. The figure demonstrates that if the building structure is

designed according to building codes, then EIP is almost constant for all four seismic

zones. But, if the design is not according to building codes, then EIP increases

dramatically with respect to the percentage of building value.

Figure 7-1: Earthquake Insurance Premium versus seismic zone (Permasteelisa 2009)

It is very important that building structures are designed based on the local code of

practice in each earthquake zone. If the structural design is not delicate enough

because of designers’ negligence or financial constraints, then earthquake insurance

premium can rise.

7.4 Cost benefits to the main structure

7.4.1 Introduction

Cost is one of the first things that come to mind when a consultant, general

contractor, or construction manager looks at the potential of additional components for

a building. The design of a movable dissipative facade is the most significant step

towards implementing a sustainable and durable façade system since it influences the

whole service life of the structure, and it has a direct impact on its construction cost. It

0

3

6

9

12

15

1 2 3 4

EIP

(%

of

buil

ding

val

ue)

Seismic zone

Acording to Code Not According to Code

238

is important to understand that there are many factors that determine the price of the

proposed system and it is important to have a system supplier on board early in the

design process. At the very least, making a call or exchanging a few written

communications with some sketches with the damper supplier to discuss up front

budgets can put to rest many fears about the cost of these systems. The overall

construction cost of a building structure is normally affected by many factors such as:

Shape & geometry

Size & regularity of floor plan

Location

Environmental forces (such as earthquake and wind) consideration in design

Market conditions

Seismic consideration

Vertical transportation strategy

Life cycle value

Structural solution

Structural and non-structural components which affect the overall cost are listed as

below:

Structure

Curtain wall (conventional)

Foundation

Mechanical and electrical equipment and other services

Each of these items does add a separate cost independent of other items. For

example, the cost of conventional curtain wall system is completely separate from the

rest and can only be affected by design cost and materials of façade panels. In this

part, parameters listed below, will be investigated financially in mid-rise and high-rise

structures in six metropolitan cities. These items for both a new and existing structure

are as below:

239

For new buildings:

Construction cost

Labour cost

Additional living area by avoiding large damper systems

Rental savings sue to extra living area available

Moreover, for existing buildings:

Retrofitting cost

Earthquake insurance premium

Sydney, Shanghai, Tokyo, Dubai, London and New York, are the selected cities,

indicated by a red dot in Figure 7-2 . Tokyo is located in high seismicity area, but

Shanghai, New York and London, are is low seismic hazard zones. Dubai and Sydney

are also selected because they are both located in medium seismic hazard zones.

Figure 7-2: Location of selected cities

Generally, there are three major approaches in measuring floor area of buildings,

which are tabulated in Table 7-6.

New York

London

Shanghai Tokyo

Sydney

Dubai

240

Table 7-6: Three major approaches in measuring floor area

Term Acronym Definition Sketch

Gross External

Area

GEA external area of a

building at each

floor level

Gross Internal

Area

GIA area of a building

measured to the

internal face of the

perimeter wall at

each floor level

Net Internal

Area

NIA Useable area within

a building measured

to the internal face

of the perimeter

walls with certain

areas excluded.

The ways of defining "floor area" depend on what elements of the building should

or should not be included, such as external walls, internal walls, corridors, lift shafts,

stairs, etc.

Table 7-7: Rent definitions

Definition Net Rent Average rent quoted per area per annum.

Gross Rent Average rent quoted per area per annum and additional costs (property taxes, service charges, operation expenses)

Cap Rate Ratio between net rent and the cost of the building or its current market value.

Gross external, gross internal and net internal areas have different definitions

according to building configuration and placement of structural columns (Cartlidge

2009). Some terminology and technical terms to be used for rent comparison and their

241

definitions are listed in Table 7-7. Construction cost, labour cost, material cost, and

rental prices for the six different cities are compared and shown in Figure 7-3 to

Figure 7-6 (J Smith and Jaggar 2006).

Figure 7-3: Construction costs

Figure 7-3 compares construction costs between the six selected cities. It can be

clearly seen that the construction cost for both office and residential building

structures in London are highest among the selected cities. The second most

expensive city in terms of construction cost is Tokyo followed by Sydney, which is in

third place. As illustrated, Shanghai is the cheapest for construction of both residential

and office buildings.

Figure 7-4: Labour costs

$0

$2,000

$4,000

$6,000

$8,000

Con

stru

ctio

n co

st (

/m2

of G

IA) Residential Office(class A)

25%33%

23% 22% 21% 19%

8%6%

10% 6% 8% 9%

17%14%

18%20% 18% 19%

50% 47% 49% 52% 53% 53%

0%

20%

40%

60%

80%

100%

Con

stru

ctio

n co

st d

istr

ibut

ion

($/m

2of

G

IA)

Others Façade Foudnation Structure

$0

$20

$40

$60

$80

$100

Sydney shanghai Tokyo Dubai London New York

Lab

our

cos

t (/h

)

General Builder Site foreman

242

Figure 7-4 compares labour costs among the six selected cities. It can be concluded

that labour costs in Sydney, London and New York are much higher than the other

three cities. Labour costs in Shanghai is the lowest among the cities.

Figure 7-5: Material costs

Figure 7-5 compares material costs among the six cities. It can clearly be observed

that the price of reinforcement in Sydney is much higher than other cities (except New

York) as labour cost in Australia is one of the highest in the world.

Figure 7-6: Rental prices and cap rate

Figure 7-6 compares rental prices and cap rates among the six cities. It can be noted

that renting an office building in Tokyo is most expensive followed by London, New

York and Sydney. Dubai has the cheapest rental prices of the six cities.

$0

$500

$1,000

$1,500

$2,000

$2,500

$3,000

Sydney shanghai Tokyo Dubai London New York

Mat

eria

l co

st (

/...)

Concrete(/m^3) Rainfrocement(/t) Façade(/m^2)

$0

$400

$800

$1,200

$1,600

$2,000

Ren

tal p

rice

(/m

2of

NIA

/yea

r) Office (class A,net rent)Office (class A,gross rent)

0

5

10

15

20

Cap

rat

e (%

)

243

7.4.2 Case studies

Both the mid-rise and high-rise structure models, which were dynamically

evaluated in Chapters 5 and 6, respectively, are going to be financially evaluated as

case studies. Once again, their specifications are listed in Table 7-8 for ease of

reference.

Table 7-8: General specifications of structural models

CS1(Mid-rise ) CS2 (High-rise) Length (m) 24 30 Width (m) 18 30 Height (m) 36 108

Number of floors 10 30 Façade (m2) 2084 13200

GIA (m2) 1860 12540

7.4.2.1 Case study 1

By using the smart façade system in outer skin of Mid-rise building, dimensions of

structural elements (beams and columns) can be reduced due to part of imparted

earthquake energy being absorbed by damper elements. Comparison between

dimensions of beam and column elements in both the conventional and smart façade

system are shown in Figure 7-7. It should be mentioned that dimensions of beams and

columns in all floors of the model are assumed to be identical in order to have easier

calculations.

(a) Conventional façade (b) Smart façade

Figure 7-7: Comparison between dimensions of beam/column elements in both the conventional and smart façade systems

700

200 400

650

200 350

244

Construction cost, construction time, labour cost, and rental income for the Mid-

rise 10-strorey structure in both conventional and smart façade systems are compared

below sections.

7.4.2.1.1 Construction cost

Construction cost for a mid-rise building structure with both convention and smart

façade systems are shown in Figure 7-8, Figure 7-9 and Table 7-9.

Figure 7-8: Building Construction cost with conventional façade and smart façade systems

As seen in Figure 7-8, by using the smart façade system, although construction of

façade panels is increased, but the total cost of the structure is reduced. The pie charts

of Figure 7-9 compares the percentage of construction costs in both conventional and

smart façade systems. As can be seen from the figure, construction cost would be

reduced by around 1% by using the smart façade system.

2.45

0.781.67

4.90

9.80

0

2

4

6

8

10

Con

stru

ctio

n co

st (

conv

enti

onal

fa

cade

,mil

lion

$)

2.28

0.73

1.78

4.90

9.70

0

2

4

6

8

10

Con

stru

ctio

n co

st (

smar

t fa

cade

,mil

lion

$)

245

Figure 7-9: Construction cost distribution with conventional and smart façade systems

Results of Figure 7-8 and Figure 7-9 are tabulated in Table 7-9 to demonstrate

financial impacts of smart façade system in a better way.

Table 7-9: Comparison of economic impacts between conventional and smart façade systems

Structure element

part

Cost (million AU$ ) Economic

impact (million $)

% building expense Economic

impact

Conventional Smart Conventional Smart % change Structure 2.45 2.28 Decrease 0.17 25 23.3 decrease 6.8

Foundation 0.78 0.73 decrease 0.05 8 7.5 decrease 6.3 Façade 1.67 1.78 increase 0.11 17 18.2 increase 6.6 Others 4.90 4.90 - 50 50 -Total 9.8 9.70 decrease 0.1 1% saving

7.4.2.1.2 Construction time

In order to calculate the total construction costs associated with labour, the previous

data must be compared with the required construction time. The construction time is

generally a complex matter, influenced by several sources of random uncertainties. The

hypotheses made in order to evaluate the financial outcome for total time and total

working costs by the input data are shown in Table 7-10 and Table 7-11.

25

817

50

Conventional façade

Structure Foundation

Façade Others

23.3

7.518.2

50

1

Smart façade

Structure Foundation Façade

Others Saving

246

Table 7-10 : Investigated parameter – Construction time

Construction time A Average labour consumption rate – formwork works wh/m2 0.65B Average formwork ratio for the entire building m2/m3 From case study C Average labour consumption rate – reinforcement

works wh/t 8

D Average reinforcement ratio for entire building t/m3 0.15E Average labor consumption rate – concrete works wh/m3 0.50F Maximum number of workers wh/hr From case study G Proportion of the average number of workers % 80 H Daily working time hr/d 8 I Concrete quantity m3 From case study J Buffer % 10 NT Nominal time wh/m3 A·B + C·D + E NT Total time days I / (F·G·H/NT) ·

(1+J/100)

It should be mentioned that consumption rates are expressed in terms of workers per

each working activity.

Table 7-11: Investigated parameter – Construction time

Construction cost

K Equipment and materials costs - formwork

$/m2 From case study

L Average formwork ratio for the entire building

m2/m3 From case study

M Equipment and materials costs - reinforcement

$/t 500

N Average reinforcement ratio for the entire building

t/m3 0.15

O Equipment and materials costs - concrete

$/m3 From case study

P Average wage $/wh From case study

Q Mark-up for overheads % 14 R Buffer % 10

NP Nominal Price $/m3 K·L + M·N + O TP Total Price $ I·(NT·P+NP)·(1+Q/100)·(1+R/100)

Based on the calculations, building construction time is decreased with smart facade

by seventy days. Figure 7-10 shows the comparison between construction time of

façade system of the 10-storey building structure in both conventional and smart

systems.

247

Figure 7-10: Comparison of construction time between conventional and smart façade systems

As an overall view, it can clearly be seen in Figure 7-11 that with similar amount of

labour, the following conclusions can be drawn when comparing smart façade system

to a conventional façade system:

• Less construction time for conventional facade

• Less labour cost for conventional facade

• Earlier rental income for conventional facade

If construction time is set to be a constant value, then less labour will be needed by

using smart façade system compared to conventional façade systems.

Figure 7-11: Construction time with constant amount of labour and time in both smart and conventional façade systems.

370

300

0

100

200

300

400

500

Conventionalfaçade

Smart façade

Con

stru

ctio

n ti

me

(day

s)

0

200

400

600

800

0 25 50 75 100

Con

stru

ctio

n ti

me

(day

s)

Labour number

Conventional façade Smart façade

Constant labours

0

200

400

600

800

0 25 50 75 100

Con

stru

ctio

n ti

me

(day

s)

Labour number

Conventional façade Smart façade

Constant time

248

7.4.2.1.3 Labour cost

As fewer employees are needed for installation of the smart façade system, then the

overall labour cost is decreased with the use of smart facade by 0.08 million dollars

which is 18.8%. Comparison of labour costs between conventional and smart façade

systems are shown in Figure 7-12 employing bar and pie charts.

Figure 7-12: Comparison of labour costs between conventional and smart façade systems

7.4.2.1.4 Rental income

As mentioned before, by using the smart façade system, the occupants can move

into the building 7 days earlier than the building with conventional façade systems.

Rental Income increases by $30,000 due to earlier occupation of the building.

Additionally, over 20 years time, rental income is increased by $190,000 due to

additional available area. Rental income increase, attained by smart façade system, is


0.65

0.52

0

0.2

0.4

0.6

0.8

Conventionalfaçade

Smart façade

Lab

our

cost

(m

illi

on$)

249

Figure 7-13: Rental income increase due to smart façade system over 20 years

7.4.2.1.5 Overall Saving

Overall saving of using smart façade system in a mid-rise builidng structure is

illustrated in Figure 7-14 for each of the selected cities. Postive values show amount

of saving and negative value illustrtate additional cost to overall system. Saving in

earlier occupation rental, construction cost, labour cost and rental income due to

additional area are considered as advantages of the proposed system and shown by

postive bars in Figure 7-14. The proposed façade system has more complexity and

obviously has higher consturction cost which is shown by negative bar in Figure 7-14.

Based on calculations, the additional cost for smart facade is $152,000. According to

the Figure 7-14 and Figure 7-15, these conclusions can be drawn as below:

Highest saving is in London with $490,000 due to expensive material and labour.

Lowest saving is in Shanghai with $152,000 due to cheaper labour and materials.

Highest additional income is in Tokyo with $350,000 due to expensive rentals.

Lowest additional income is in New York, Dubai & Shanghai with $170,000 due to

cheaper rentals.

0.19

0.03

0

0.05

0.1

0.15

0.2

Additional Area Earlier Occupation

Ren

tal

(mil

lion

$)

250

Figure 7-14: Comparison of building component expenses profit by using smart façade system

Figure 7-15: Comparison of overall saving by using smart façade system

For better understating, the results in Figure 7-14 are tabulated in Table 7-12.

Table 7-12: Values (million$) of building component savings by using the smart façade system

City

Additional cost of smart

façade Labour cost

Earlier occupation

rental income

Additional area rental income (20

years)

Construction cost

Sydney -0.145 0.116 0.03 0.145 0.290Shanghai -0.145 0.007 0.01 0.145 0.145

Tokyo -0.145 0.058 0.04 0.290 0.290Dubai -0.145 0.015 0.01 0.145 0.145

London -0.145 0.101 0.03 0.290 0.435 New York -0.145 0.135 0.01 0.145 0.145

-$0.5

-$0.4

-$0.3

-$0.2

-$0.1

$0.0

$0.1

$0.2

$0.3

$0.4

$0.5

Sydney Shanghai Tokyo Dubai London New York

Sav

ing

Additional cost of smart façade Labour costEarlier occupation rental Additional area rental income (20 years)Construction cost

0.44

0.16

0.49

0.13

0.62

0.29

$0.0

$0.2

$0.4

$0.6

$0.8

$1.0

Sydney Shanghai Tokyo Dubai London New York

(mil

lion

$)

251

7.4.2.2 Case study 2

By using the smart façade system in outer skin of the building, dimensions of

structural elements (beams and columns) can be reduced due to part of imparted

earthquake energy is absorbed by damper elements. Comparison between dimensions

of beam and column elements in both the conventional and smart façade systems are


(a) Conventional façade (b) Smart façade

Figure 7-16: Comparison between dimensions of beam/column elements in both the conventional and smart façade systems

Construction cost, construction time, labour cost and rental income for the high-rise

30-strorey structure with both conventional and smart façade systems are compared as

below:

7.4.2.2.1 Construction cost

Construction cost for a mid-rise building structure with both convention and smart

façade systems are shown in Figure 7-17, Figure 7-18 and Table 7-13.

90 100

700

700

800

900

252

Figure 7-17: Building Construction cost with conventional façade and smart façade systems

As seen in Figure 7-17, by using the smart façade system, although construction cost

of façade panels is increased, but total the cost of the structure is reduced.

Figure 7-18: Construction cost distribution of a mid-rise building with conventional and smart façade systems

The pie charts of Figure 7-18 compare the percentage of construction cost using

both conventional and smart façade systems. As can be seen from the figure,

construction cost would be reduced by around 0.8% by using the smart façade system.

Results of Figure 7-17 and Figure 7-18 are tabulated in Table 7-13 to demonstrate the

financial benefits of the smart façade system.

46.55

14.9431.61

92.95

186.04

0

40

80

120

160

200

Con

stru

ctio

n co

st (

conv

enti

onal

fa

cade

,mil

lion

$)

44.81

14.3632.48

92.95

184.59

0

40

80

120

160

200

Con

stru

ctio

n co

st (

smar

tl

faca

de,m

illi

on$)

25

817

50

Conventional façade

Structure Foundation

Façade Others

24.1

7.717.5

50.0

0.8

Smart façade

Structure Foundation Façade

Others Saving

253

Table 7-13: Comparison of economic benefits of conventional versus smart façade systems

Structure element

part

Cost (million AU$ ) Economic

benefits (million $)

% building cost Economic benefits

Conventional Smart Conventional Smart % change Structure 46.5 44.8 decrease 1.74 25 24.1 decrease 6.9

Foundation 14.9 14.3 decrease 0.59 8 7.7 decrease 6.9 Façade 31.6 32.4 increase 0.87 17 17.5 increase 7.2 Others 92.9 92.9 - 50 50 steadyTotal 186.0 184.5 decrease 1.45 0.8% Saving

7.4.2.2.2 Construction time

Based on the calculations, building construction time is decreased with the use of

smart facade system by 68 days. Figure 7-19 shows the comparison between

construction time of façade system of the 30-storety building structure using both

conventional and smart systems.

Figure 7-19: Comparison of construction time between conventional and smart façade systems

7.4.2.2.3 Labour cost

As fewer employees are needed for installation of the smart façade system, then

overall labour cost is decreased with smart facade by 1.16 million dollars, which is

15.7%. Comparison of labour cost between conventional and smart façade systems are

shown in Figure 7-20 using bar and pie charts.

446

378

0

100

200

300

400

500

Conventionalfaçade

Smart façade

Con

stru

ctio

n ti

me

(day

s)

254

Figure 7-20: Comparison of labour cost between conventional and smart façade systems

7.4.2.2.4 Rental income

As mentioned before, by using the smart façade system as an outer skin of building

structure, the tenants can occupy the building 68 days earlier than when using

conventional façade systems. Rental Income increases by $4.79M due to earlier

occupancy of the building. Additionally, over 20 years, the rental income is increased

by $1.6M due to additional rentable area. Rental additional income attained by smart

façade system, is shown in Figure 7-21.

7.83

6.67

0

2

4

6

8

10

Conventionalfaçade

Smart façade

Lab

our

cost

(m

illi

on$)

255

Figure 7-21: Additional rental income due to using smart façade system

7.4.2.2.5 Overall Profit

The overall savings using smart façade in a mid-rise builidng structure is illustrated

in Figure 7-22 for each of the selected cities. Postive values show amount of savings

and negative values illustrtate additional cost to overall system. Saving in earlier

rental, construction cost, labour cost and rental income due to additional letable area

are considered as advantages of the proposed system and shown by postive bars in

Figure 7-22. The proposed façade system has more complexity and obviously has

higher consturction cost which is shown by negative bars in Figure 7-22. Based on

calculateions, additional cost when employing smart facade is $0.15M. According to

the Figure 7-22 and Figure 7-23 the following conclusions may be drawn:

Highest saving is in London ($0.5M) due to more expensive material and labour.

Lowest saving is in Shanghai ($0.15M) due to cheaper labour and materials.

Highest additional income is in Tokyo ($0.31M) due to more expensive rentals.

Lowest additional income is in New York, Dubai & Shanghai ($0.17M) due to

cheaper rentals.

1.60

4.79

0

1

2

3

4

5

Additional area Earlier occupancy

Ren

tal

(mil

lion

$)

256

Figure 7-22: Comparison of building component expenses/savings by using smart façade system

Figure 7-23: Comparison of overall savings when using smart façade system

For better understating, the results of Figure 7-22 are tabulated in Table 7-14.

Table 7-14: Values (million$) of building component expenses/savingsby using smart façade system

Cities

Additional cost of

smart façade

Labour cost

Earlier rental

occupancy

Additional area rentalincome (20

years)

Construction cost

Sydney -0.87 1.16 4.79 1.595 1.16Shanghai -0.87 0.073 3.63 1.16 0.58

Tokyo -0.87 0.58 6.82 2.32 1.305Dubai -0.87 0.145 2.90 1.015 0.58

London -0.87 1.02 6.09 2.03 1.885New York -0.87 1.305 2.90 1.015 0.725

-$2

$0

$2

$4

$6

$8

$10

Sydney Shangai Tokyo Dubai London New York

Pro

fit

Additional cost of smart façade Labour costEarlier occupation rental Additional area rental income (20 years)Construction cost

7.83

4.50

10.01

3.63

10.01

4.93

$0.0

$2.0

$4.0

$6.0

$8.0

$10.0

$12.0

Sydney Shangai Tokyo Dubai London New York

Sav

ing

(mil

lion

$)

257

7.5 Strategies and approaches

Any project requirements such as seismic loads, wind loads, live load deflections,

acoustics, and any impact or blast resistance criteria need to be considered and

analysed before any budget pricing takes place. The system needs to be designed

specifically on a case-by-case basis and be tailored to individual project requirements.

Therefore, not all projects use the same fittings. They are customized as required, but

work of a similar design concept. Without discussing the project with a structural

glass vendor, it is entirely possible that the design proposal will not be possible as

envisioned. The façade engineering team needs to review the test reports from the

damper system manufacturers to ensure these systems meet or exceed the design

criteria for a specific project. If these reports are not satisfactory or readily available,

the team needs to require that specific tests be performed in relation to the

specifications. This will ensure building owners will receive the highest quality

product to limit potential liabilities down the road.

7.6 Summary and conclusions

The design principles for the smart facade system were proven through extensive

numerical analysis. The proposed system like any other material will degrade and lose

their functionality over time. Despite the fact that, deterioration of façade materials is

an unavoidable phenomenon, but service life of the damper material can be extended

to an optimum value by considering possible durability issues at the stages of design,

construction, operation, maintenance and repair. These stages are discussed briefly

below:

258

1. Design:

Design of the system should be in a way that steel layers of the damper system

have minimum contact to open air and surrounding environment. Based on various

climates and regions in which the damper is going to be installed, different strategies

need to be considered at the design stage to avoid any deterioration that may be caused

by environmental factors. The designer must focus on detailing the system in such a

way to minimize the probability of premature degradation of the components. Finally,

effects of such factors as resisting structural and environmental loads, heat and air

transfer, preventing water and moisture infiltration, and acoustics should be

considered in façade design.

2. Construction

The damper layers need to be ordered from a reputable and experienced company,

for example the 3M company which is one of the few companies in the world that can

make the delicate viscoelastic damper layers. Construction and assembly of the

damper system would be done in a workshop for lower quantities or in a factory for

mass production depending on budget allocated to the project. For this reason,

supervision by a specialist is quite essential during construction activities to ensure

that the installation work is conducted in accordance with the original design and

specifications. Depending on types of structure, namely, as existing or a new building

structure, a careful attention needs to be paid during the installation process of the

façade damper systems. High standard of workmanship is needed to minimize the

issues arising from poor installation of the panels.

259

3. Operation and maintenance

The higher is the quality of the layers, the more durable is the system. For that

reason, regular quality control inspections is paramount during service life of the

system to ensure that performance of the system is in line with the original design and

specifications. Inspections need to be done by fully trained technicians under the

supervision of façade/material experts. The manager should keep a record of the

results of all inspections throughout the service life of the damper system and should

assess the safety of street-level facilities by referring to these records. Any

deteriorating salts or ions need to be removed promptly to enable the assessors to

perform a better visual inspection. If serious deteriorations take place before the end

of the service life, and no maintenance work is conducted to correct the deterioration,

the system may lose its functionality and cause serious issues. Corrective maintenance

would be needed, only if a detailed inspection reveals that rehabilitation work is

needed. Therefore, planned maintenance must be performed to ensure safety and

serviceability of the whole system.

4. Repair:

In the absence of regular maintenance, the façade bracket elements would continue

to deteriorate, and finally its repair/ rehabilitation could become expensive, requiring

replacement of the damper system. A specialist needs to consultant with the designer

and decide the best rehabilitation process for restoring the bracket damper to its

original condition. Replacement of components is necessary at the end of their service

life to ensure maximum dynamic performance. It should be said that, the module

width and height, number of support points, glass make-up (thickness and

performance) required, back-up structure, are all important factors which depend on

260

needs of the client and would change in each smart façade system contract. System

pricing can vary greatly depending on the requirements. The facade inspector should

be experienced in the field of stability, defect deterioration, forensic investigation,

remedial engineering mechanisms and expert witness relating to specific materials and

facade assembly.

The design team can advise building owners of potential cost saving measures by

modifications in design if there is a budget range in mind. Post-earthquake repair costs

and repair time are two crucial parameters, which need more research. Because,

detachment of façade components from the building may lead to social and economic

issues, as well as increase in injury or death to the occupants and pedestrians. Smart

façade systems, if designed and installed properly, possess a better economy of scale

on a cost per square meter basis.

Using the smart façade system has the following benefits and advantages.

1. Additional rentable area

Rental area losses are when using large scale dampers but there would be no loss in

the rental area when smart façade system is used to enhance dynamic response of

the main structure.

2. windows or/and doors

No closing or opening process is needed in smart façade panels.

3. Time

No temporary loss of rental income happens when using the smart façade system,

but tenants need to move out of the offices when other retrofit techniques are

selected.

4. Re-design costs

261

No structural modification is needed to the building structure when smart façade

system is selected as a method of retrofit.

262

CHAPTER 8

CONCLUSION AND FUTURE

WORK

263

In this chapter, the considerable results and conclusions of the research presented in

this dissertation are outlined. The outcomes of this research are utilized to form

recommendations for practical design of the proposed bracket system. Moreover, an

experimental test plan to investigate the three-dimensional behaviour of a corner

cladding system is proposed as future work.

8.1 General conclusions

The main purpose of the research discussed in this dissertation was to study the

effect that movable dissipative cladding system (energy absorbing damping device)

has on the multistorey building structures seismic response. These systems can be

effective and are considered in order to mitigate the seismic effect on various

structural building systems when the dominant seismic frequencies are close to the

natural frequencies of the main structure. The feasibility of the proposed system was

proven by numerous computer analyses. In this thesis, analytical models were created

in ANSYS APDL to study the effects that movable cladding systems have on the

seismic response of multistorey building structures.

A number of different structural models incorporated with energy absorbing

connections and under different earthquake records were modelled to attain a

comprehensive understanding of the efficiency of the proposed damper connections.

A 10-storey and a 30-storey building with facade systems, in both 2D and 3D with and

without energy absorbing connections, were investigated under four different

benchmark earthquake records suggested by the International Association for

Structural Control and Monitoring. It is necessary to mention that the results of the

comprehensive analysis on 2D structural models are not included in this thesis in order

to avoid undue increase in the size of this thesis. In addition, the results obtained from

264

2D studies were consistent and a subset of those presented in the thesis for 3D

buildings. Each of the studied building facade systems with energy absorbing

connections behaved in a different manner, and the effectiveness of the energy

absorbing connections varied under different earthquake motions. This can be

attributed to the varying intensity and frequency content of the applied earthquakes.

The effects of the cladding on the seismic response of the frame structure were

evaluated by performing modal analyses and dynamic time-history analyses of the

analytical models.

For the mid-rise building structure, the fundamental period of the frame structure

with conventional façade panels was 0.98 seconds, while the fundamental period of

these models with movable cladding was 1.04 seconds. Thus, the movable cladding

systems increased the fundamental period of the models by a maximum of 5.7%,

compared to the frame structure with fixed façade panels. For the high-rise building

structure, the fundamental period of frame structure with conventional façade panels

was 2.82 seconds, while the fundamental period of these models with movable

cladding was 2.92.

Thus, the movable cladding systems increased the fundamental period of the

models by a maximum of 3.5%, compared to the frame structure with fixed façade

panels. It was noticed that, the mode shapes and effective modal mass percentages

were also not significantly affected by the movable facade systems. Time-history

analyses were performed to evaluate the nonlinear dynamic response of the frames

excited by ground motions with a wide range of intensities. Four ground motions were

selected to perform the time-history analyses. The parameters recorded during the

analyses included top floor displacement, maximum inter-storey drifts, maximum

265

floor accelerations, maximum plastic hinge rotations, and base shear force. The time-

history results were plotted in several combinations. Detailed analytical models of the

buildings with cladding systems were shown in Chapters 5 and 6. The force-

deformation relationship of the cladding connectors were obtained from experience

and prior studies of the industry partner, Permasteelisa company. The results showed

that the integration of the dissipative damper connection to the building facade

systems enhanced the energy absorption effectiveness and decreased the seismic effect

on all levels of the structure.

Additionally, façade connection properties had significant influence on seismic

response of the primary building structure. The closer investigation of the structural

buildings showed that a reduction in the axial stiffness (parallel to applied earthquake)

of the proposed bracket system did not have much influence in controlling the seismic

response of the main structure. However, reduction in shear stiffness with the

optimum value has shown to have a major impact in reducing response in all seismic

response parameters. Beyond this optimum value, the seismic response of the main

structure, started to increase. From the results, it is also evident that incorporation of

dissipative damper system in the facade system played an important role in altering

drift between stories of the main structure. Results have shown that the connection

stiffness and energy absorption capacity have a great influence in mitigating the

adverse effects of earthquakes. To conclude, the main findings of the study are

presented here:

1. It is feasible to use energy-absorbing connections in building facade system to

control response of main structure by dissipating part of the imparted seismic

energy. Viscoelastic dampers have proven to be very efficient for this purpose and

266

the connection properties have significant influence on the response. They display

optimum values of stiffness and damping based on intensity and kind of

earthquake record.

2. The energy absorbing connections placed in direction perpendicular to applied

earthquake force were able to control the deformation and forces in structural

elements with reasonable differential displacement between the frame and the

façade.

3. The influence of façade mass on the seismic mitigation was investigated as

well. Results showed that increase in the mass ratio resulted in higher reductions

in relative building response. However, the increase in the mass ratio is not

economical.

4. Seismic mitigation of the building system response was possible when the

natural frequencies of the structure were within the range of dominant frequencies

of the earthquakes.

5. The damping connections with shear deformation in the majority of cases were

able to produce remarkably high improvement and reduced the seismic effect on

the building structure at all levels. The results showed that the best performance

of VE damper connections in most cases was observed to be achieved in the

upper storeys in comparison to the lower and middle storeys. As noted before,

placement of the system in the top storey levels has reduced the magnitude of the

measured response parameters.

6. In some cases, in both mid-rise and high-rise building structures, in lower

storey levels and under the applied earthquake excitations, the VE damper

connections were not as effective as top floors and an increase in the magnitude

of the response parameters were noted. This can be explained by the fact that

some of the applied earthquakes have an unusually low dominant frequency range

in some parts of their excitation.

7. Bracket elements with value of shear stiffness between 10-90 N/mm can be

effective in reducing the response of the main structure in both mid-rise and high-

rise building structures. As it is impossible to install bracket elements with

different values of shear stiffness because of complexity of their design, then

implementing a semi-active or active control system is needed for the proposed

concept.

Overall, the use of the comprehensive time-history analyses in Chapter 5 and 6

coupled with financial feasibility and assessment of the proposed system in Chapter 7

provide a rational framework for selecting appropriate stiffness for designing the

cladding in various geographical areas.

8.1.1 Application and contribution of this research to design

Double-skin Facades have made a rapid dispersal into the commercial markets such

as Australia, North America and specially the Middle East. In seismic analyses and

design, structural engineers have typically disregarded the extra stiffness and damping

that the cladding system may add, which could be beneficial to the building’s seismic

performance. This study has indicated the possibility of developing new façade

connections with appropriate properties to reduce the response of the main structure

during earthquakes. The idea of advanced cladding connections developed in this

research came from the wish to take advantage of energy dissipation due to relative

movement of the cladding panels and structural frame.

However, these systems necessitate cladding systems to encounter significant

lateral movements to create any promising effects; therefore, crucial criteria such as

267

268

the appearance, water tightness, and air tightness due to the relative panel-structure

movement could be threatening to the concept. On the other hand, earthquake

excitations are not as common as wind forces which impact the structure more

frequently. Therefore, except for any possible damage to the façade system, reduction

in structural response and consequently decrease in rate of casualties, will be

welcomed by engineers, building owners and the community. From the several time-

history analyses carried out here, it was evident that with the implementation of

appropriate connection properties, the seismic response of the main structure can be

considerably reduced. Moreover, the connection deformation and the connection

forces can be kept within reasonable and practical limits by applying pre-defined load

deformation behaviour to them. Engineers, often, would prefer to optimize a system

for a particular property such as low energy consumption. Particularly in the design of

buildings, this is difficult because of conflicting requirements such as optimizing

daylight and minimizing solar gain.

Thus, co-optimizing is essential. The relationship between architects, engineers,

and facility managers must be managed carefully to develop and use advanced control

features that do not overwhelm users or the lack of management may render the

advanced facades inefficient. Additional design tools must be developed so that

structural and façade designers may easily investigate effects of movable double-skin

facades on seismic performance of the structure during design stage. It is noteworthy

to mention that, acceptance of the proposed façade system is linked to the additional

expense to the building owner as well as the architectural and environmental benefits

indicated.

269

8.2 Recommendations for future research

The research discussed in this dissertation has focused to answer the important

questions related to the seismic capability of movable cladding systems, especially

double skin façade systems, in multistorey buildings. However, more research in this

field will provide more comprehensive results, in terms of structural geometries,

cladding configurations, and connection types, it also assists code and standard

committees to revise the code and add a section for designing movable dissipative

façade systems in earthquake prone zone areas. The data used for these analyses are

based on analytical and very limited experimental tests on movable cladding system

components. Thus, to improve the performance of the proposed system, more

experimental tests on existing and new cladding systems, will provide statistical data

needed for design of various types of cladding connectors typically found in

construction.

In addition, more discussions with cladding manufacturers and contractors will

provide additional data. The physical tests will provide not only validate the

performance of the proposed system, but also provide valuable data to re-calibrate the

properties of the damper material in ANSYS models. Determining the repair costs and

repair time of the proposed cladding system are also necessary parts of future

experimental work in order to fully evaluate the practicality of the system. More

research on the structural response of non-regular structures with different height and

plan irregularities subjected to three-dimensional (horizontal and vertical) earthquake

excitation is critical for completing this journey.

270

8.2.1 Further research that would improve and complement this thesis

It has been observed from the contents of this research that the focus of this PhD is

on numerical modelling and verification of the proposed concept. Further tasks need

to be considered which are listed as below:

a. Façade panel distortion, local and general deformation in connections need to be

looked at

b. Micro modelling of the proposed attachment which is shown in Figure 8-1 is

needed. The modelling needs to be sketched in Solid Works program and

exported to ANSYS or ABAQUS for further and accurate analysis.

Figure 8-1: Details of proposed connection for attachment of façade outer skin to slab of main structure

271

c. Gap width between outer and inner façade layers need to be optimized. The

random variable consists of length of axial damper element and needs to be

evaluated numerically and experimentally.

8.2.2 Proposed experimental test program

The major part of the research in this dissertation focused on computer simulations

of cladding systems and parametric studies. To gain additional insight into the seismic

performance of multistorey buildings with movable cladding system, an experimental

test needs to be performed. The proposed experimental testing program will provide

insight into the three-dimensional behaviour of the movable cladding systems. This

section outlines a proposed testing program to evaluate the response of a full-scale

portion of a movable cladding system.

The goals of the tests are to understand how the cladding system components

interact with the main structure and understand their dynamic behaviour. Additionally,

their interaction together as a uniform layer is very crucial to avoid possible collision

and internal damage and needs to be monitored carefully. The results of the tests

should provide some validation to the analytical results presented in this dissertation.

The simulation needs to provide insight into how panels and connectors behave during

an earthquake along the entire height of the building and to validate the analytical

models developed in ANSYS APDL. Locations of possible damage and identification

of the failure modes of the connections need to be determined as well.

8.2.2.1 Test setup, specimen design and terminology

The building will be indicative of mid-rise, moment resisting frame structures

common in commercial real estate. Although the material of the building structure was

reinforced concrete in the numerical analysis, but in order to have an easier set up and

installation, a steel structural frame can be built with pinned connections so that the

frame has less lateral resistance. The beams and columns are considered to be box

sections. A sketch of the south-west elevation of a corner specimen is illustrated in

Figure 8.2. Two panels, which are in red colour on top level of building structure, are

selected for experimental test analysis. They can be attached to the main frame

individually or ideally together to evaluate their interaction with primary structure and

with each other. Details of the proposed experimental test model in different angels

are shown in Figure 8-2 to Figure 8-7.

Figure 8-2: South-west sketch of the building structure and elevation of the specimen

Although the test specimen does not represent all kinds of cladding panels for the

mid-rise and the high-rise structural buildings, but the full-scale corner specimen

provides a unique opportunity to evaluate the interaction between the movable facade

panels as the frame moves. Corner subassembly experiences largest inter-storey drift

and largest post-yield drifts during seismic loading. Additionally, dynamic response

and panel interaction at the corners of the building structure are difficult to understand

272

273

and it would be a good opportunity to evaluate this interaction. The façade system,

connection types, and connection locations need to be considered the same as those

discussed in this dissertation. Full-scale cladding assemblies measuring 3600mm

(one storey) tall by 1500mm wide should be tested to investigate the interaction of the

cladding panels, in plane and out of plane. Two quasi-statically tested specimens are

expected to provide information about the overall behaviour of the system,

connector’s behaviour, the interaction of façade panels at the corners, and the

progression of damage in dissipative damper device.

Figure 8-3: Sketch of details of experimental model

274



275



More details of the proposed damper system and its configuration in building

structure are shown in Figure 8-8 and Figure 8-9.

276

Figure 8-8: Side views of attachment of the proposed damper system to slab of primary structure

277

Figure 8-9: Top view of attachment of damper system to slab of primary structure

278

APPENDIX A

SECTIONS OF THE

STRUCTURAL MODELS

279

Midrise Structural Model

Beam Levels Beam section Equivalent Section

1-3

4-6

7-9

280

Beam Levels Beam section Equivalent Section

10

281

Column Levels Column section Equivalent Section

1

2-3

282


4

5

6

283


7-8

9-10

284

Highrise Structural Model

Beam Level Beam section Equivalent Section

1-10

11-17

18-30

285

Column Level Column section Equivalent Section

1-9

10-16

286

Column Level Column section Equivalent Section

17-19

20-23

24-30

287

APPENDIX B

THESIS TERMINOLOGY

288

1. Movable concept: the idea to define/allow the outer skin of DSF system to move

back and forth in both in-plane and out-of-plane directions during applied earthquake.

2. Optimum dynamic properties for façade brackets: refer to dynamic behaviourand energy dissipation capabilities of connections during applied earthquake. Optimum properties have more effects on reduction of dynamic response of structure building during applied earthquake.

3. Advanced connectors: Bracket system which has capability to dissipate part ofapplied load by moving back and forth in both in-plane and out-of-plane directions during applied earthquake.

4. Elastic Conventional façade bracket: Structure with an elastic behaviour inwhich dimensions of structural elements are designed and calculated so that only minor damage is sustained during earthquake excitation. The façade system used in this model has fix(conventional) brackets.

5. Plastic Conventional façade bracket: Structure system with elements (beam &column) which has plastic behaviour at both ends. Dimensions of structural elements are designed and calculated so that damage occurs mainly in beam elements during earthquake excitation. The façade system used in this model has fixed (conventional) brackets.

6. Conventional bracket: or rigid bracket or fixed connector is an element whichhas mechanical properties of (modulus of elasticity, strength) stainless steel. The element is only allowed to move few millimetre which is similar to what is recommended in cladding industry.

7. Bracket: element which connects façade panel to main structure or secondarysteel support system.

8. Façade frame: is referred to curtain wall frame. Façade frame is the outercovering of a building in which the outer walls are non-structural, utilized to keep the weather out and the occupants in.

9. Façade column: the vertical elements mainly made of aluminium and are alsonamed mullion.

10. Main façade frame: steel support for the outer layer of double skin façadesystem.

11. Fix façade: Façade system which is connected to the main structure withconventional (fixed or rigid) bracket system.

12. Flexible façade: façade system which is connected to the main structure withadvanced connectors.

289

13. Low axial stiffness: façade bracket elements which can move out of plane (backand forth) due to force

14. Damper device: a dissipative element which is replaced by the conventionalbracket and connect the façade system to the main structure.

15. Plastic joints: introducing a bracket system which is capable of having plasticdeformation. Plastic joints forms in the bracket element to dissipate part of the applied forces.

290

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