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Engineering Failure Analysis 17 (2010) 213–225

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Engineering Failure Analysis

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Collapse of an industrial steel shed: A case study for basic errorsin computational structural engineering and control procedures

A. Brencich *

DICAT – Department of Civil, Environmental and Architectural Engineering, University of Genoa, Via Montallegro 1, 16145 Genova, Italy

a r t i c l e i n f o

Article history:Received 20 April 2009Accepted 7 June 2009Available online 21 June 2009

Keywords:Steel trussComputational structural analysisStructural testing

1350-6307/$ - see front matter � 2009 Elsevier Ltddoi:10.1016/j.engfailanal.2009.06.015

* Tel.: +39 0103532512; fax: +39 0103532534.E-mail address: [email protected]

a b s t r a c t

Collapses that do not produce other than economic losses remain often unknown either tothe public and to the community of engineers. The technical causes of any collapse deserveattention even if no scientific research is needed since they often show that design andcontrol procedures and code provisions may fail in preventing errors both in the designand in the building phase of standard structural engineering. In this paper the collapseof an industrial steel shed, under a 10 cm layer of fresh snow, is discussed showing thatits collapse resulted from a chain of errors, in the design phase, during its assemblageand in the final inspection and control phase.

� 2009 Elsevier Ltd. All rights reserved.

1. Introduction

The history of structural collapses often follows the ‘‘pendulum principle”: when a collapse has a significant impact on thepublic opinion, economic and academic resources are earmarked for investigating its causes and preventing further similarevents. After new code specifications are issued and, hopefully, new design and inspection procedures are introduced andadopted by the technical community, the memory of the collapse slowly fades till a new collapse, of similar or different type,gives new excitation to the pendulum [1–6] in 1940, 1967 and 1981, respectively, are outstanding examples of this principle.

When collapses only cause economic losses, which may happen in the case of industrial facilities, the limited impact, if any,on the public opinion gives no excitation to the pendulum and no resource is earmarked for studying these cases. Nevertheless,these ‘‘minor” collapses may provide useful suggestions either for scientific research and for technical and professional practice.

This paper deals with the collapse of a steel industrial shed resulting from a chain of errors: (i) severe mistakes in thecomputerized structural analysis of the shed; (ii) lack of any control of the numerical results; (iii) lack of controls in thebuilding phase; (iv) superficiality in analysing the warnings provided by the structure during the assembly phase; (v) mis-understanding of the results of a load test, performed at the end of the construction which, showing severe structural defi-ciencies, was considered completely satisfactory. This case study outlines at least a couple of issues: (i) a chain of errors isneeded for a structure to collapse; (ii) also severe code provisions, that should have been applied to the specific case, may notprevent the disaster if they are considered as formal requirements emptied of any technical meaning. Due to privacy reasons,nothing is said on the location of the structure.

2. Background

The shed, built in 1993–1994, was divided into two separate blocks 80 m long and 70 m wide, Fig. 1, for a global dimen-sion of 160 m � 70 m. Each block consisted of a series of seven parallel main truss girders (allignments 1–7) connected by

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Fig. 1. General plan of the shed (one block only), internal trusses, end frames and secondary trusses.

214 A. Brencich / Engineering Failure Analysis 17 (2010) 213–225

secondary truss beams (allignments a–m) and supported by two HE 500 B steel columns/truss 10.25 m high (allignments band l). The end frames were different in their central part and supported by six additional HE 200 B columns, Fig. 1. The trusswas 3.55 m high at midspan (1/14th of the span) and 2.50 m at the free edge of the cantilever (1/4th of the cantilever span);the standard horizontal length of the truss units was 2 m but for the two central units that were 3 m long. The elements ofthe truss where all twin angles with dimensions ranging from 80 � 80 mm to 150 � 150 mm and thickness in-between 10and 16 mm. The steel was S355 and connections made use of u 24 mm, 6.8 class bolts [7].

In the following, the equivalence between European and American standards is provided:

HE 500 B: I shaped, height: 500 mm, width: 300 mm, A = 2.39 � 105 mm2, J = 1.07176 � 109 mm4 similar to US W18 � 119.HE 200 B: I shaped, height: 200 mm, width: 200 mm, A = 5.38 � 103 mm2, J = 5.696 � 107 mm4, similar to US W8 � 31.Angles 80 � 80 � 10 mm and 150 � 150 � 16 mm correspond to US L21/2 � 21/2 � 1/2 and L6 � 6 � 5/8, respectively.S355 steel [7] equivalent approximately to ASTM A572/572 M Grade 345 carbon steel [8].

The design wind + snow loads were assumed as large as 0.6 kN/m2, coherent with the design code of the early 90’s, whichare nowadays exceeded by the new local building code, asking for that area a reference wind + snow load close to 1 kN/m2.

For assembly reasons, the girders where pre-assembled into six pieces jointed (through assemblage joints) on two col-umns, at the locations outlined by the two dashed lines of Fig. 1 and at midspan.

The structural shape of the main trusses presents an untypical feature: it is somewhere in-between a Warren and a Moh-nie beam with both compressed and tensile braces; being the shed in an open flat area, and being the technical equipment ofthe shed limited to the lighting, the reasons for such a choice are unknown; the designer did not change this choice originallymade by the customer. It is worth noting that the truss was a statically determined structure.

3. The collapse

The shed collapsed in late February 2001 at lunch-time under the load of 8–10 cm of fresh snow (estimated weight:0.2 kN/m2), Fig. 2, at an average temperature close to 0 �C. No one was injured and only economic losses were suffered.

All the main trusses showed a specific collapse mechanism, Fig. 2a, related to the assemblage joints (dashed sections ofFig. 1): the northbound (towards the mountains) joint collapsed due to fracture of the connecting plate, Fig. 3.

Fig. 2. The shed some days after the collapse, when part of the cantilevers had already been demolished. (a) Front view (note the aerator in the central partof the roof), north side on the right, south on the left. (b) Lateral view (partial).

A. Brencich / Engineering Failure Analysis 17 (2010) 213–225 215

4. The design phase

The investigations carried out after the collapse found a series of errors, the first of which originated in the design phase,that were not corrected because of many reasons, mainly because of the hurry in building the shed and because of technicalsuperficiality in dealing with the construction and its final verification. In the following, these errors are discussed in detail.

The structural analysis was computerized and related to one of the internal main trusses only. In order to reduce the com-puting time, only half of the truss (full truss can be recognized in Fig. 1) was analysed, on the basis of symmetry properties,according to the scheme of Fig. 4a (circles mark the assemblage joints of the truss). This choice prevented the designer fromconsidering asymmetric loads, probably assuming that the reduced slope of the pitched roof could cause a uniform distribu-tion of the snow loads. No specific attention was paid to wind, which is, on the contrary, one of the most severe loads in thatspecific area close to the coast. The design wind/snow loads, as already mentioned in Section 2, were 0.6 kN/m2; the aeratorat midspan, Fig. 2a, was not taken into account.

The external constraint (point A of Fig. 4) was the 10.25 m long HE 500 B steel column (maximum slenderness in the trussplane: 97) and was represented by the designer as a perfect hinge restraining both vertical and horizontal displacements.This made the full truss to be over constrained in the axial direction, which corresponds to the case of perfectly rigid col-umns, but does not fit the actual restraint of a 10.25 m long steel column.

Assuming a homogeneous distribution of the live loads (i.e. 0.6 kN/m2), the horizontal reaction force at point A was esti-mated to be as large as 676 kN (67.6 tons) and the distribution of axial thrust like in Fig. 4b, with bold lines referring to com-pressed struts and thin lines to tensile elements. According to this distribution of internal forces, Section 3 had a 249 kN axialthrust; such a low value is probably the reason for choosing Section 3 as the assemblage joint.

Fig. 3. Typical collapse of a shed – schematic.

Fig. 4. (a) Structural scheme adopted in the design phase for the truss girder. Distribution of the axial force according to: (b) the model of Fig. 4a; (c) aproper scheme; (d) modelling also the steel column.


No control had been performed on the automatic solution, which is somehow not easy for the braces (but for the cantilevers)due to the mixed structural scheme of the truss in-between a Warren and a Mohnie truss, as already discussed. For the upper andlower chords, an easy verification of the automatic analysis could be performed taking into account that the bending moment ofa beam is sustained by a truss girder by a compressive force in the upper chord and a tensile one in the lower chord.

If we compare the axial thrust of the chords close to midspan according to the structural scheme adopted in design,Fig. 4b, we recognize that the axial thrust in the upper chord is almost three times higher than in the lower chord, whichbecomes suspicious noting that the difference between the two values is exactly the horizontal reaction in section A. Anotherwarning was given by the large difference in the compressive thrust of the lower chord in the elements close to the restraintsA: �264 kN on the left, �940 kN on the right (the difference is again the horizontal reaction force at point A). Additional con-


trols could have been performed also by means of Ritter sections to verify the axial thrust here and there, at least in theassembly joints, since the truss girder is a statically determined structure.

The aforementioned untypical distributions of axial thrust are due to the horizontal overstraining of the truss: hinge Aapplied to the truss an horizontal force of 674 kN which provided a compressive contribution to the axial thrust in the lowerchord. Since much of the lower chord is in traction, such an additional force is underconservative for a large part of the truss.

Due to the reduced stiffness of the steel column, the horizontal displacements of point A should be considered free, thusassuming only a vertical restraint at point A. In this case, the distribution of axial thrust turns out to be substantially differentfor the lower chord, Fig. 4c. Table 1 summarizes the axial thrust at Sections 1–6 (Fig. 4a) for the ”design” scheme of Fig. 4b,for the ‘‘correct” scheme of Fig. 4c, and for an additional model that substitutes the external restraint A with the actual steelcolumn, Fig. 4d.

As already discussed, two significant loads had not been taken into account: the aerator at midspan (1.5 kN/m) and aneccentric snow load close to the aerator (in the following: a triangle 1.5 m � 1 m against the aerator along the whole shed);for the letter load, the full truss girder needs to be modelled. Table 1, last two lines, shows the design axial thrust that shouldbe used if these two load cases were taken into account, pointing out that these two ‘‘neglected” loads would increase theaxial thrust in the critical sections some 20%.

One of the connection joints of the truss was located in Section 3 just where the ‘‘design” scheme of Fig. 4b showed avalue of the axial thrust in the lower chord of 249 kN and was designed with two bolts (two shear sections/each) on eachside of the joint only, Fig. 5. The same joint in the upper chord, with a �832 kN thrust, was designed with seven bolts (fourshear sections/each), which should had been considered as a suspicious design feature.

Table 1Axial thrust in Sections 1–6 (Fig. 4a) for different structural schemes.

Model Axial thrust at section (Fig. 4a) [kN]

1 2 3 4 5 6

(1) Axially over constrained – Fig. 4b +422 �1098 +249 �832 �940 10(2) Simply supported – Fig. 4c +1098 �1098 +925 �832 �264 10(3) With steel columns – Fig. 4d +1098 �1098 +913 �832 �273 10(4) Simply supported – Fig. 4c + aerator +1226 �1241 +1013 �908 �264 14(5) Simply supported full truss + aerator + eccentric snow load +1345 �1355 +1086 �972 �264 16

Bold figures refer to the assemblage joint in the lower chord.

Fig. 5. Connection joints of Sections 3 and 4: 7 bolts (four shear sections/each) in the upper chord (Section 4) and 2 bolts (two shear sections/each) in thelower chord (Section 3).


The actual value of the axial thrust in the assemblage joint (Section 3, Fig. 4) should have been estimated as large as925 kN for a uniform snow load, and 1086 kN if the aerator and an eccentric snow load were considered, which is some3.7 and 4.4 times the value assumed for the design of the joint. These comparisons show that the safety coefficients implic-itly included in the design strengths had been completely vanished by the error in the external restraint of the truss girder,Fig. 4b.

On the basis of the characteristic values for the ultimate strength of bolts (480 N/mm2 for 6.8 bolts) and steel (inter-nal plate and L profile, 355 N/mm2 for S355 steel), the assemblage joint would sustain the axial thrust summarized inTable 2. It is clear that the weak link is the internal plate, even though the actual values at collapse are to be expectedhigher than those in Table 2 since characteristic values are lower than the average values to be found in practice. What-ever the actual ultimate value, the values of Table 2 show that the assemblage joint was clearly a weak point of thetruss with inhomogeneous strength of its components. The comparison between the expected characteristic ultimate ax-ial force in the joint (580 kN, Table 2) and the expected design value (1086 kN, Table 1, case 5) shows dramatically thedeficiency in the design phase due to the tricky error in the external restraint. Further comments on this issue are givenin Section 7.

5. The construction phase

The end frames, Fig. 1, reproduce the standard internal girder in the cantilevers but are substantially different in the inter-nal part: no brace was used assuming a frame response of the girder. During construction, when just the roof plates werebeing placed on something less than half of the shed, the extreme part of the cantilever showed a vertical displacementof approximately 150 mm (from the technical documents of the contractor). The building company thought that such a largedisplacement was due to the tolerance between the bolts and the holes. Detailing of the chord-column connections wasrather poor and was typical of a hinged connection unable of transferring bending moment, Fig. 6; therefore the quads inthe frame actually behaved something like a hinged quad. No additional structural analysis was performed to understandthe origin of such a huge displacement of the cantilever end, so that the pathologic scheme of the end frames was not iden-tified. New braces were introduced in the quads and the 150 mm displacement was compensated in some way that could notbe identified later on.

Table 2Ultimate axial load for the assemblage joint for different limit states.

Limit state Bolts shear collapse Internal plate tensile collapse L profile tensile collapse

Ultimate axial load (kN) 870 580 1490

Fig. 6. End frames: (a) original design; (b) modified by introduction of diagonal braces.


A couple of comments need to be outlined:

(i) A 150 mm displacement is not a ‘‘small displacement” and should have warned all the people involved in the con-struction; the bolt–hole tolerance would account for a displacement that would be 10 times smaller than the observedvalue.

(ii) The additional braces changed the mechanical response of the structure from a frame to a truss; for this reason, thiskind of strengthening should not be performed if detailed and very careful structural analyses are not performed.

6. The load test

A load test was performed on the structure as soon as its structural components had been completed; such a control wasnot compulsory but strongly recommended by the local building code.

The control Authority asked the contractor just to ‘‘apply loads to a single truss equivalent to the design wind + snow loadincreased some 10%” and to measure the most relevant displacements; a specialized sub-contractor was charged to do theload test. No estimate either of the expected displacements and of the axial thrust in the members was provided before thetest, so that it was performed as a blind test.

The loads were applied to truss n 3, which was approximately in the centre of the shed, keeping the secondary trusses,Fig. 1, perfectly effective. In this way, the loads were applied to one truss but were sustained by several trusses due to thetransversal distribution of the loads in the shed truss structure. It is worthwhile noting that the main trusses were 3.00 mhigh on the average and the secondary trusses were 2.70 m high, so that their contribution to the overall resisting structurecould not be neglected a priori.

Two load distributions (load cases) were used:

(i) Load case 1: load distribution of Fig. 7a, load cycles to zero with increasing peak load: 10–20–30–40–43 kN/jack.(ii) Load case 2: only the internal nodes loaded, Fig. 7b, 47 kN/jack applied without any preliminary cycle. The load could

not exceed 47 kN/jack since the counterweights of the jacks where 50 kN concrete blocks.

The last two load cycles for the load distribution 1, Fig. 7a, (4th and 5th runs) and for the second one, Fig. 7b, (one cycleonly, 6th run) are shown in Fig. 8 referring to the midspan. No measurement was performed on the trusses close to the testedone so it was impossible to record the transversal distribution of the applied load.

The load test showed unusual and unexpected outcomes:

(i) 4th load cycle, up to 40 kN/jack: residual displacement equal to 41% of the maximum value; loading part of the dia-gram not linear, Fig. 8a.

(ii) 5th cycle, up to 43 kN/jack: residual displacement equal to 7% of the peak value; loading part of the diagram not linear,Fig. 8b; significant hysteretic shape of the cycle.

(iii) 6th load cycle, up to 47 kN/jack: load distribution of Fig. 7b; irrecoverable displacement equal to 16% of the peakvalue; loading part of the diagram not linear, Fig. 8c; hysteretic shape of the diagram.

The unrecovered displacements were attributed to the sliding of the bolts in their holes due to the hole-bolt tolerance.The residual displacements of the last two cycles (7% and 16%, the latter with a different load distribution) where assumed

Fig. 7. Load test and distribution of the loading devices: (a) load case n 1; (b) load case n 2. Load distribution for (a) the first 5 cycles (maximum load:430 kN); (b) last run (maximum load: 470 kN).

Load case 1 - run n. 4

0

10

20

30

40

50

0 5 10 15 20 25 30 35

Displacement at midspan [mm]

Load

/ Ja

ck [k

N]


0

10

20

30

40

50

0 5 10 15 20 25 30 35

Displacement at midspan[mm]

Load

/ Ja

ck [k

N]


0

10

20

30

40

50

0 5 10 15 20 25 30 35 40 45 50

Displacement at midspan[mm]

Load

/ Ja

ck [k

N]

(a) (b)

(c)

Fig. 8. Load–displacement response for the two load cases: (a) load case 1, run 4; (b) load case 1, run 5; (c) load case 2, run 6.


as corroborating such an interpretation: the largest part of this displacement was believed to be contained in the 4th run,which was neither included in the final report nor referred to in the text of the final report in order to raise no discussion.No attention was paid to the hysteretic shape of the cycles.

In fact, sliding of the bolts due to the hole-bolt tolerance would be responsible for irrecoverable displacements that arelower than the residual values measured in the tests of Fig. 8. The large hysteretic response and the repeated large value ofthe unrecovered displacements should have provided a serious warning, raising serious objections on the interpretation gi-ven to the test data.

A careful analysis of the data would have suggested different conclusions.

(i) Run 4: peak load of 40 kN/jack and 41% of unrecovered displacement.(ii) Truss completely unloaded.

(iii) Run 5: peak load of 43 kN/jack and 7% of unrecovered displacement.

these facts could be typical of yielding, as the hysteretic shape of the cycle should have suggested.It is worth noting that the load tests, that were intended to be studied only on the basis of the unrecovered displacements

at the end of the last cycles, had not been designed, i.e. the maximum expected displacements had not been computed, sothat no in real time control could be performed during the load test.

Experimental tests on the connection joint of the lower chord, Fig. 3, performed in laboratory after the collapse, showedthat the actual yield axial force (for concentric axial loading) of the joint was 560 kN, with ultimate load carrying capacity of840 kN, with collapse due to the fracture of the internal plate of the joint. These values are some 45% higher (840 kN against580 kN) than the values estimated in Table 2 assuming the code provisions (ultimate strength 355 N/mm2), which are char-acteristic and not actual values.

At the peak load of cycles 5 and 6 the axial thrust in the assemblage joint, provided by an elastic FEM model of the wholethree-dimensional structure of the shed, Fig. 9, taking into account the actual transversal distribution of the applied loads,Table 3, was estimated as large as 650 kN and 770 kN, respectively. Comparing these values with the measured yield andultimate values we obtain a clear explanation of the unrecovered displacements showed in Fig. 8: they were due to yieldingof the internal plate of the assemblage joint. A preliminary design of the load tests could have provided a serious warning onthe measured displacements.

A very rough estimate of the expected displacement at the peak load was performed by the sub-contractor, after the loadtest, assuming the simplified model for the truss girder of a 1-dimensional beam with bending stiffness related to the upperand lower chords only. The shear stiffness was assumed unbounded and not taken into account; the whole of the testing load

Fig. 9. Complete 3-D fem elastic model of the shed.

Table 3Transversal distribution of the load at load cycles 5 and 6 foreseen by the model of Fig. 9.

Loaded truss (Fig. 1)

1 2 3 4 5

Load case 1 – run 5Displacement at midspan (mm) 1.6 18.3 31.6 20.3 7.1Participation factor (%) 2 23 40 26 9

Load case 2 – run 6Displacement at midspan (mm) 0.5 12.7 23.0 14.1 8.2Participation factor (%) 1 23 42 26 8


was applied to this model. In this way, the peak displacements (at midspan) were estimated to be 43.5 mm for the 5th cycleand 57.2 mm for the 6th one. Being the maximum measured displacements 27.8 (64% of the estimated value) and 42.5 mm(74% of the calculated value) respectively, the load test was assumed to be satisfactory and the shed was opened to its finalusage.

More accurate computations would have foreseen completely different results, either because of the excessive load givento the model and because the shear contribution to the global displacements was neglected. Table 4 summarizes the resultsprovided by more accurate analyses assuming two models: (i) a complete 2-D model of the truss, taking into account boththe bending and shear deformability of the truss, no transversal distribution of the applied load taken into account; (ii) a 3-Dcomplete model of the shed, representing the estimated response of the ‘‘real” structure.

The complete 3-D model of the shed takes into account the transversal distribution of the loads due to the secondarytrusses and shows an expected displacement, for elastic behaviour, that is significantly lower than the measured value. Itis only a wrong estimate of the load sustained by the loaded truss that makes theoretical displacements much higher thanthe measured ones. The correct displacements, if computed in advance, would have warned that something different fromthe elastic response (yielding somewhere in the girder) was happening. But it was thought that a load test is satisfactoryprovided that the measured displacements are below the estimated ones; the lower the measured values the better thestructural response. Such an approach is clearly unjustified and the consequences of this rather diffused opinion are wellrepresented by the collapse of this shed.

Taking into account the actual load distribution on the loaded truss, the test load would be found to be as large as 0.26 kN/m2, which is less than half the design required value and 40% of the required test load (0.60 kN/m2 + 10%).

On the basis of the previous discussion, we can now conclude that the load tests led the tested truss to yielding in the 4th,5th and 6th cycle. Therefore, the ‘‘decreasing” values of the unrecovered displacements (41% of peak value at cycle 4, 7% atcycle 5, 16% at cycle 6) are due to the hardening of the assemblage joint. The collapse during the load test did not take placeonly because the adjacent girders helped the loaded one to sustain the loads.

Table 4Displacements of the loaded truss: measured values and expected ones.

Displacement at midspan (mm)

Measured Estimated by the sub-contractor

S-C/measured

2-D truss model with full appliedload

2-D/measured

3-D completemodel

3-D/measured

Load case5

27.8 43.5 1.56 54.1 1.95 23.0 0.83

Load case6

42.5 57.2 1.35 77.4 1.82 31.6 0.74


7. Why did the shed collapse 7 years after its construction?

The design and assemblage errors discussed in the previous sections identify only a part of the causes of collapse since afact remains obscure: the collapse (2001) took place 7 years after the construction of the shed (1994) and the estimated loadat collapse (10–15 cm of fresh snow, estimated 0.20–0.30 kN/m2) was not exceptional. Since such a load was probably notnew to the structure, why did it cause the collapse in 2001 and not before?

During the assembling phase no control was performed not only on the design and assemblage procedures but also on thematerials, even though the local building code asked the materials, of any type of profile used in the trusses, to be charac-terized by means of laboratory tests (tensile yield stress, tensile strength and Charpy toughness, at least three tests for eachtype of profile) on specimens directly derived from the profiles actually used for the structure. Since also the joint platesshould have been tested, the total number of tests required, and not performed, to identify the materials was as large as36. These tests were performed after the collapse to investigate the material quality and gave answer to the fundamentalquestion of this paragraph.

The assemblage joints collapsed in two different ways: due to brittle fracture of the internal plate, Fig. 10, the vast major-ity, due to ductile fracture of the internal plate some of the collapsed joints, Fig. 11. In each truss, the assemblage joint of oneside only (northbounds) collapsed, while the opposite joint was found severely damaged but not fractured, Fig. 12. The rea-son for this, and the answer to the basic question of this paragraph, come from material testing of the internal plates, Table 5;while the L profiles showed the material to be the design one, the internal plates of the assemblage joints provided interest-ing information.

Two assemblage joints that exhibited brittle and ductile collapse, respectively, were chosen to identify the material of theinternal plates by means of standard tensile and Charpy tests [9] on the undamaged parts of the joints; Table 5 shows theresults. The plate exhibiting ductile fracture was actually the required S355 grade C steel, while the other plate, that col-lapsed due to a brittle fracture, clearly exhibited unusually low toughness values and too low values for the yield stressand for the ultimate strength. In order to better understand the low toughness values given by the brittle plate, the ‘‘undam-aged” parts of that plate, far from the fracture, were used to obtain further nine Charpy specimens to be used for artificialageing in the following way:

Group A: Three specimens were subjected to thermal treatment [10] in order to recall them back to their original ‘‘brandnew” condition;Group B: Three specimens were not treated, in order to test the material as it was at the moment of the collapse;

Fig. 10. (a) and (b) Brittle fracture two assemblage joints after collapse; (c) detail.

Fig. 11. (a) Ductile fracture of an assemblage joint; (b) detail.

Fig. 12. Yielded (but not collapsed) assemblage joint.

Table 5Mechanical properties of the internal plates of assemblage joints exhibiting brittle and ductile fracture. Toughness measured at 20 �C.

fy (N/mm2) fu (N/mm2) Kv Kv – av. Kv – min

Plate with brittle fracture (Fig. 10) 323 394 39, 27, 5 24 5Plate with ductile fracture (Fig. 11) parallel to the chord axis 425 548 88, 103, 60 93 60Plate with ductile fracture (Fig. 11) normal to the chord axis 401 552 – – –


Group C: Three specimens were subjected to thermal treatment and artificial ageing [10,11]: (i) a 5% ± 0.5% compressiveplastic strain was induced in the rolling direction; (ii) heating till 250 �C for 30 min and than cooling in air at roomtemperature.

The measured toughness is summarized in Table 6 where it appears clear that: (i) the material is very susceptible to age-ing; (ii) taking into account the data of Table 5, the steel of some of the plates of the assemblage joints was not the requiredS355 grade C steel; the strength properties could suggest a S275 steel but the toughness values underline some unidentifiedsteel. The reduction of toughness (approximately half the ‘‘brand new” value in 7 years, suggests some kind of steel for carconstruction, which is produced in the same city in which the shad had been built).

Table 6Charpy Kv toughness measured at 20 �C for the steel of internal plates of the assemblage joints collapsed with a brittle fracture.

Kv Kv – av. Kv – min

Group A – brand new 104, 102, 113 106 102Group B – as it was at collapse = brand new + 7/8 years 51, 77, 80 69 51Group C – after artificial ageing 3, 3, 4 3 3


The reasons for plates in the assemblage joints of a different and unknown steel could not be understood. Looking at thedocuments of the building period, one finds that all the pieces were delivered to the construction site in 1993 and assem-blage, due to procedural problems, was started more than 12 months later; in this period some plate could have gone lostand could have been replaced with the unknown material, but this is only a conjecture. Due to limited funding, tests couldnot be performed on all the assemblage nodes so that the percentage of the inadequate plates is not known.

Data of Tables 5 and 6 show that in February 2001 the shed was not the same as in 1994, when it was just finished, sincesome of the plates of assemblage joints (and possibly of some other joint) had already changed their mechanical propertiesdue to premature ageing. Since the unknown steel has been proved to reduce its toughness in a short period of time, we canfind the reason for the collapse in 2001 in the premature ageing and in the environmental condition in the day of collapse. Inthat day the snowfall was followed by strong winds that reduced the temperature to 0 �C downtown, and possibly belowzero in the construction site, which was close to the sea in a flat open area.

Besides, Fig. 1 shows that the collapsed assemblage joints were almost entirely on the northbound side of the shed, sothat it can be conjectured that collapse was activated by the snow gathered against the aerator on the northbound sideof the shed.

The combination of premature aging, low temperature, additional loads not considered in the design (aerator at midspan),eccentric load (snow against the aerator) and wind blows (which accounts for dynamic impulsive loading of the structure),along with the already discussed errors in the design phase are all the causes that lead the shed to its collapse the 27th ofFebruary 2001.

8. Other issues

Other circumstances add to the series of errors discussed in the previous sections and, since all of them could haveavoided the collapse, are worthwhile discussion.

(i) The design documents contained no verification of the bolted connections. This rises a doubt: had the connections everbeen verified?

(ii) The lack of verification of the bolted connections was found only after the collapse. Since several engineers were incharge of the construction (the designer, the engineer who followed the assemblage; the engineer in charge of the finalverifications; the chief engineer of the building company at least) and had to deal with the project, it seems thatnobody paid any attention to the design documents.

(iii) The computations (related to the structural design) were asked to the designer some months after the shed had beencompleted, not before starting construction, as the local codes (and proper engineering practice) asked.

(iv) Detailing of the truss girder was poor, leaving several aspects, also of structural relevance, unsolved.(v) The 70 m long trusses were connected to the HE 500 B columns by 4 u16 mm bolts, which shows another weak point

of the shed.(vi) The aerator, Fig. 2, was assembled at midspan in the last phases of the construction; nobody controlled whether this

device had been taken into account in the design phase. Also this ‘‘negligible” load was ‘‘discovered” only after thecollapse.

All these issues share a common feature: if attention had been focused on, at least, one of these aspects, the weakness ofthe whole project and construction procedure would have appeared, providing an occasion for a general verification of thestructure. Even if we cannot be sure that the collapse could be avoided, the probability of discovering the deficiencies wouldhave greatly increased.

9. Discussion

The collapse of the steel shed discussed in this paper belongs to the large series of ‘‘minor collapses” that produce onlyeconomic losses and that remain unknown to the public at large. Therefore, the investigation on the causes of the collapse isgiven no circulation also in the engineering community. As it happens quite often, also in this case the causes of collapse donot outline any new problem for Structural Engineering but only a long series of human errors and completely ignored con-trol procedures: for these reasons, the causes of this collapse are potentially present in the vast field of Civil Engineering andput forward the need for further discussion on this issue.


How to avoid this kind of collapses is the crucial issue deriving from the case study discussed in the paper. Could strictcontrol procedures avoid the chains of errors that had been performed in this case? The local regulations and building codeare amongst the most rigorous in the industrial countries but had no effect in this case, where all the fulfilments were eithersatisfied only from a formal point of view or completely ignored.

Large part of the job is for the academics and their teaching efforts to make clear to the new engineers that StructuralEngineering should never be obscured by the fog of routine, where controls become ‘‘a series of papers”.

Further results could be obtained if the control Authority were independent from the contractor and from the purchaser.In the case of this shed, it was the contractor who could choose the ‘‘independent” authority that had to control the buildingphase and the purchaser could choose the engineer in charge of the final verifications, which is probably one of the weakestpoints of a control procedure.

References

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[10] UNI ISO 5446-84. Ferromanganese. Specification and conditions of delivery. Italian version of the equivalent ISO 5446/80 (in Italian).[11] UNI EN 10001-94. Definition and classification of pig-irons. Italian version of the equivalent EN 10001-94 (in Italian).