ROCK SOCKETS IN SOFT ROCK

By George H. Glos, III,1 and Oliver H. Briggs, Jr.,2 Members, ASCE

ABSTRACT: This paper describes the results of two full scale rock socket load tests conducted in conjunction with the foundation design of a sulfur dioxide removal plant located at the Four Corners Generating Station near Farmington, New Mexico. The tests were conducted in a soft shaley sandstone with an av­erage unconfined compressive strength less than 1,500 psi (10.3 x 103 kPa). The strength parameters of the sandstone were investigated by means of laboratory triaxial shear tests as well as Menard pressure testing in the field. The sockets were loaded to a maximum of 2,000 kip (9.2 x 105 kg) and were monitored by sets of Carlson strain gages along the length of the socket and along the bottom diameter. The results of these tests indicate the skin resistance mobilized to be nonuniform and that socket length greater than 2 diam or 3 diam do not sig­nificantly affect socket capacity:

INTRODUCTION

Caissons socketed in rock are frequently used to support heavy loads. "< The recent expansion of industrial growth in the sun belt of the United States where geologic conditions often favor this foundation type has greatly increased their use. The high cost associated with a load test of a caisson results in economic justification only on rare occasions. i

Until recently, rock sockets were designed based on a uniform stress distribution derived from empirical data. Recent studies based on elastic '.'•] theory indicate that mobilized skin resistance varies from the top to the ' bottom of the socket. Rosenberg and Journeaux proposed a design pro­cedure which takes this into account. Pells and Turner developed design charts based on this nonuniform stress distribution. Ladanyi and Kout- •,' softas provided limited field data which would tend to support this y concept.

Two socket load tests were performed on caissons socketed into soft "' rock in conjunction with the foundation design of a sulfur dioxide re- :";? moval plant located at the Four Corners Generating Station near Far- ua

mington, New Mexico. This paper presents the results of these tests. :"

'.•.3

SUBSURFACE CONDITIONS ,

The rock formations encountered in the southwest are often consid- = erably softer than their midwestern and eastern counterparts. Two rock ,5

'Geotechnical Engr., Arizona Public Service Co., Phoenix, Ariz. Supervisor, Civ.-Struct. Design, Arizona Public Service Co., Phoenix, Ariz. Note.—Discussion open until September 1, 1983. To extend the closing date

one month, a written request must be filed with the ASCE Manager of Technical and Professional Publications. The manuscript for this paper was submitted for review and possible publication on June 23,1982. This paper is part of the Journal of Geotechnical Engineering, Vol. 109, No. 4, April, 1983. ©ASCE, ISSN 0733-9410/83/0004-0525/$01.00. Paper No. 17922.

525

J. Geotech. Engrg. 1983.109:525-535.

Dow

nloa

ded

from

asc

elib

rary

.org

by

IZM

IR Y

UK

SEK

TE

KN

OL

OJI

EN

STIT

USU

on

03/0

7/15

. Cop

yrig

ht A

SCE

. For

per

sona

l use

onl

y; a

ll ri

ghts

res

erve

d.

Downloaded from http://www.elearnica.ir

sockets were constructed in the Picture Cliffs formation. The formation generally consists of an interbedded sandstone and shale. The water ta­ble was located at a depth of approximately 8 ft (2.4 m). The two sockets were located at depths of 50 ft and 46 ft (15.24 m-14.02 m), and were designated east and west sockets, respectively. Both sockets were lo­cated primarily in a shaley sandstone. The piers were excavated down to the test interval and then cased off and bailed out, allowing work to proceed in the dry. The rock in the test sections had moisture contents generally less than 4%. Menard Pressure meter tests performed at var­ious locations in the formation indicated an average projected limit pres­sure of 327 kip/sq ft (15,656 kPa). The deformation modulus ranged from 124 ksi-330 ksi (855 MPa-2,275 MPa) with an average of 260 ksi (1,792 MPa). Core recovery averaged 98% with a rock quality designation of 80%.

The east socket was located in a horizontally bedded shaley sandstone with some coal stringers. Core recovery (NQ) was 98%, and the RQD was 88% in the test socket section. Four unconfined compressive strength tests were performed on core samples from the socket section. Uncon­fined compressive strengths ranged from 639 psi-1,994 psi (4,405 kPa-13,746 kPa) with an average of 1,343 psi (9,258 kPa). Schmidt rock ham­mer tests were performed vertically on the lip of the rock socket. Values ranged from 16-21 with an average of 18.

The west socket was also located in a horizontally bedded shaley sand­stone. The core recovery was 95%, with an RQD of 74%. Unconfined compressive strengths ranged from 667 psi-1,710 psi (4,598 kPa-11,788 kPa) with an average of 1,213 psi (8,362 kPa). Schmidt rock hammer tests were also performed vertically on the lip of this socket. These ranged from 8-22 with an average of 16.

A review of this data indicates the east socket to be slightly more com­petent of the two. The range of Schmidt Rebound Numbers would cor­respond to a Rock Modulus of 1,000 ksi-8,000 ksi (6,894 MPa-55,152 MPa). The RQD values indicate a range in rock modulus of 600 ksi to 2,250 ksi (4,136 MPa-15,511 MPa), based on results of large plate load tests per­formed by Coon and Merritt.

Caisson Installation and Set-Up.—Caissons 42 in. (1.07 m) in diam were advanced to the socket elevation. Upon reaching the socket ele­vation, a 42-in. diameter (1.07-m) casing was placed. This allowed the caisson to be pumped out and the socket to be excavated in the dry. The sockets were then advanced using 24-in. (0.61-m) diam auger. Two sets of grooves 4 in. (100 mm) high and 3 in. (76 mm) deep were con­structed in each socket in order to increase the side friction, and give a larger apparent diameter. The exact geometry of each socket is shown in Fig. 1.

The lip of each socket was hand-cleaned, and a styrofoam doughnut was placed on it in order to prevent bearing from occuring there. A 30-in. (762-mm) diameter casing was then placed on the styrofoam. The annular space between the inner and outer casing was left void. Six Carl­son elastic wire strain meters were attached to the reinforcement bar in each socket. The strain gages were located 14 in. (356 mm) below the top lip and 1 in. (25 mm) above the bottom of each socket. The location of strain gages is also shown in Fig. 1.

526

J. Geotech. Engrg. 1983.109:525-535.

Dow

nloa

ded

from

asc

elib

rary

.org

by

IZM

IR Y

UK

SEK

TE

KN

OL

OJI

EN

STIT

USU

on

03/0

7/15

. Cop

yrig

ht A

SCE

. For

per

sona

l use

onl

y; a

ll ri

ghts

res

erve

d.

East Socket Wast Socket

FIG. 1.—Geometry of East and West Sockets

Reaction was provided by four rock anchors cemented in place with the top of the cemented anchor located 9.75 ft (2.97 m) below the bottom of the socket and 10 ft (3.05 m) to each side. Analysis of the anchorage system indicates that their presence increased the apparent settlement by approx 5%, based on the method proposed by Poulos and Davis.

Theoretical Estimates of Ultimate Bearing Capacity.—Full-scale load tests in sedimentary rock evaluated by Williams and Pells determined both peak and residual side shear, the results of which indicate that for most sandstones and shales side shear ranges from about 7%-35% of the unconfined compressive strength. The mean and upper limit values were chosen because it is felt that these more nearly reflect the confined compressive strength, the ultimate side shear in the range of 40 ksf to 60 ksf (1,915 kPa-2,872 kPa) is calculated.

End bearing values based on the Griffith Fracture Theory yield ulti­mate capacities of 1,250 kip/sq ft-2,500 kip/sq ft (59,850 kPa-119,700 kPa). These are again based on mean and upper limit unconfined com­pressive strengths.

Pressure meter tests have been found to underestimate generally rock stiffness. Based on a limit pressure of 327 kip/sq ft (15,656 kPa), the ultimate bearing pressure is calculated to be 1,470 kip/sq ft (70,383 kPa), and the ultimate side shear is 12 kip/sq ft (575 kPa).

TEST RESULTS

The load settlement curves for the east and west sockets are shown on Figs. 2 and 3, respectively. The curves show the total and net de­flection of the socket as well as the calculated elastic compression of the pier. The elastic modulus of the concrete was estimated from unconfined compressive tests. Loads were generally held for 30 min. Movements were measured by linear voltage displacement transducers capable of measuring to the nearest 0.001 in. (0.025 mm). The instrumentation was

527

J. Geotech. Engrg. 1983.109:525-535.

Dow

nloa

ded

from

asc

elib

rary

.org

by

IZM

IR Y

UK

SEK

TE

KN

OL

OJI

EN

STIT

USU

on

03/0

7/15

. Cop

yrig

ht A

SCE

. For

per

sona

l use

onl

y; a

ll ri

ghts

res

erve

d.

LOAD SETTLEMENT CURVES

East Socket

1000 1500 Load In Kips

FIG. 2.—Load Settlement Curve—East Socket

LOAD SETTLEMENT CURVES

West Socke t

> 0.3

| 0.3

z °-4

0.5

0.6

0.7

I elct pomi

la t & r e s

d e l ion

/

^

t o t

is t lo

/

BOtt

il dc

lemfi

f l e e

nt

Hon

500 1000 1500 2000 Load In Kips

FIG. 3.—Load Settlement Curve—West Socket

calibrated for temperature effects by reading during a 30-hr period with no load.

The shear collars (Fig. 1) may have caused the effective diameter to increase from 2.0 ft-2.5 ft (610 mm-760 mm). Fig. 4 shows the calculated relationship between diameter and Rock Modulus at a load of 1,000 kip (4,448 kN) based on the method proposed by Pells and Turner. The be­havior of the rock sockets was reasonably elastic in nature, as indicated by the almost complete strain recovery of the east socket to the 1,260 kip (5,605 kN) load; however, the load versus deflection curves were less than straight. The permanent deflection at this point was then 0.07 in. (1.8 mm). The reloading curve matched the primary loading curve to within 0.01 in. (0.25 mm). At the completion of the tests, the total per­manent deflections were 0.14 in. (3.56 mm) and 0.18 in. (4.57 mm) in

528

J. Geotech. Engrg. 1983.109:525-535.

Dow

nloa

ded

from

asc

elib

rary

.org

by

IZM

IR Y

UK

SEK

TE

KN

OL

OJI

EN

STIT

USU

on

03/0

7/15

. Cop

yrig

ht A

SCE

. For

per

sona

l use

onl

y; a

ll ri

ghts

res

erve

d.

-*

ck

Mo

du

lus

ec

SETTLEMENT VS ROCK MODULUS

SOO-

4 0 0 -

300-

200-

100-

0-c

SOCKET DIAMETERS OF 2.0 & 2.S

SOCKET LENGTH: S.O APPLIED LOAD : 1000 KIPS

I \ \ SOCKET DIAMETER : 2.0

S 0 C K E T ^ = ^ ^ = = ^ ^ DIAMETER : 2 . 5 ~ ^ = = = = = = = = = = = _

.1 .2 .3 .4 .5

Calculated Settlement - Inches

FIG. 4.—Settlement Versus Rock Modulus

the east and west sockets, respectively. Figure 5 shows the change in Rock Modulus with settlement. The data

is based on the method proposed by Pells and Turner. The equation used to calculate the rock modulus based on the measured settlement is

P e = x Ip

ERa y (1)

Rock Modulus Er IN Ksl

ELASTIC SETTLEMENT OF ROCK SOCKET

<

0.1

Inch

es

r I 0.

Cal

cula

t

g

< 3 U

r

/ w

3

>o to

i IS

/ c

^ S § !

c

I I j i

L :!(i i( (S

3 J

: d6

' 'MT III K

l l |T

S

/

<e

c c c

(

I i

l:5C

• 1

0

3 < 3 > K

-700

o

o

T

c e c c

3 1

3 1

IP

2 . 0

1.0 0.9 O.S 0.7 0.6 0.5 0.4 0.3

0 . 2

0 . 1

I

-

r 7 / :

A

" N:

I

1 ' IP1

*&5>

E p

4-o

f 2

0 . 5 1

2

^ T ~ — — 1 - ' ) 0

I I I

JL

_

-

--

-

0 2 4 6 8 10

L/a

Alter Pells & Turner

FIG. 5.—Calculated Settlement versus FIG. 6.—Elastic Settlement of a Rock Rock Modulus Socket

529

J. Geotech. Engrg. 1983.109:525-535.

Dow

nloa

ded

from

asc

elib

rary

.org

by

IZM

IR Y

UK

SEK

TE

KN

OL

OJI

EN

STIT

USU

on

03/0

7/15

. Cop

yrig

ht A

SCE

. For

per

sona

l use

onl

y; a

ll ri

ghts

res

erve

d.

PERCENT OF LOAD TRANSFERRED TO SOCKET BASE

FIG. 7.—Settlement versus Load Transferred to Socket Base

in which e = settlement; P = applied load; a = socket radius; ER = Rock Modulus; and Ip = the settlement influence factor based on socket ge­ometry and the ratio of moduli between the pier and the rock (Fig. 6). Fig. 5 has been corrected for the effect of embedment and the anchorage system.

Figure 7 compares percent of load carried by end bearing based on Pells and Turner's theory with that actually measured. The difference is believed caused by the highly stratified nature of the rock, and the prob­ability of rock modulus increasing below the socket base. This may have caused a greater portion of the load to be transferred to the base.

Load Transfer Data.—The load transfer data are based upon the pier instrumentation. All instruments produced readings which are consid­ered to be consistent and correct. However, instrument 6 in each socket

EAST SOCKET

top of socket

SOO 1000 1S0O 2000

Vertical Load - Klpe

0

» 1

f 2

©

% 3 o 0

w 4

WEST SOCKET

top of socket

/ /

/ /

/ , I /

I 1 /

' *

/

X

500 1000 1500 2000

V Brtlcal Load - KEpe

FIG. 8.—-Soeket Depth versus Vertical Load—East Socket

FIG. 9.—Socket Depth versus Vertical Load—West Soeket

530

J. Geotech. Engrg. 1983.109:525-535.

Dow

nloa

ded

from

asc

elib

rary

.org

by

IZM

IR Y

UK

SEK

TE

KN

OL

OJI

EN

STIT

USU

on

03/0

7/15

. Cop

yrig

ht A

SCE

. For

per

sona

l use

onl

y; a

ll ri

ghts

res

erve

d.

Sond va Vertical Load

gee*

SOO 1000 1S00 2000

Vertical Load Kips

FIG. 10.—Bond versus Vertical Load

produced data considerably lower than their counterparts located on the rebar cage. (The data from these instruments are excluded from this sec­tion, and are considered under "End Bearing.")

Figures 8 and 9 show the measured load distribution within the east and west sockets, respectively. These figures show the measured de­crease in load, with increasing depth into the socket. It can be seen from the figures that below a socket length of approx 2 diam nearly all load is transferred into end bearing and bond becomes minimal. This effect is generally predicted from elastic theory if no slip is assumed.

Figure 10 shows the bond mobilized vs. the applied load. The top two curves represent the bond mobilized above the top set of strain gages. The lower curve represents the average bond over the entire socket. The bond mobilized at the top instrument elevation was approx 79 kip/sq ft and 89 kip/sq ft (3,782 kPa and 4,260 kPa) in the east and west sockets, respectively. It is apparent from this figure that these values are very close to the ultimate value and are beginning to peak out. The average bond mobilized in each socket was approx 45 kip/sq ft or 312 psi (2,151 kPa). This roughly corresponds 45% and 25% of the unconfined com­pressive strength of the rock for the maximum and average bond mobilized.

laat Socket West Socket

a 300-

Applled Load

Socket Base Inat. No. Socket Base lnst.No.

FIG. 11.—-End Bearing—East Socket FIG. 12.—End Bearing—West Socket

531

J. Geotech. Engrg. 1983.109:525-535.

Dow

nloa

ded

from

asc

elib

rary

.org

by

IZM

IR Y

UK

SEK

TE

KN

OL

OJI

EN

STIT

USU

on

03/0

7/15

. Cop

yrig

ht A

SCE

. For

per

sona

l use

onl

y; a

ll ri

ghts

res

erve

d.

f 1

600 1000 1500 2000

Vertical Load - KSF

FIG. 13.-— End learing versus Vertical Load

End Bearing.—The load distribution in end bearing is shown in Figs. 11 and 12 for the east and west sockets, respectively. The curves gen­erally conform to what would be expected of a rigid footing resting on an elastic medium. The curves also indicate some apparent eccentricity which was also reflected in the upper instrumentation values.

Figure 13 shows the measured end bearing for the instruments located on the rebar cage. This figure indicates that at higher loads, the end bearing began to increase disproportionately with respect to the bond mobilized. This is because the bond had reached or nearly reached its ultimate capacity and was behaving in a more nearly plastic rather than elastic manner.

Creep.—Several loads were held for as long as 12 hr during the course of the test. The greatest amount of movement occurred during 12 hr hold at 2,000 kip (8,896 kN) on the west socket. This movement was approx 0.07 in. (1.8 mm). This translates to 0.009 in. (0.25 mm) per log cycle of time.

At low load increments, creep was extremely small being on the order of 0.002 in. (0.05 mm) per log cycle of time at a 645 kip (2,869 kN) load (west socket).

The rock sockets were tested approx 30 days after pouring, and con­crete creep was probably a significant, although unquantitied, portion of the total creep in these tests. The total magnitude of the creep due both to concrete and rock, however, was indicated as being quite small and not of practical significance from the design standpoint. Accord­ingly, it has not been plotted for this paper.

CONCLUSIONS

The results of the load tests appear to lead to the following conclusions:

1. Published correlations of either the Schmidt Rebound Numbers or

532

J. Geotech. Engrg. 1983.109:525-535.

Dow

nloa

ded

from

asc

elib

rary

.org

by

IZM

IR Y

UK

SEK

TE

KN

OL

OJI

EN

STIT

USU

on

03/0

7/15

. Cop

yrig

ht A

SCE

. For

per

sona

l use

onl

y; a

ll ri

ghts

res

erve

d.

the RQD values with Rock Modulus were not accurate in this soft rock. 2. Theoretical estimates based on empirical data tend to underesti­

mate greatly the bond capacity of rock sockets in this test. 3. The rock sockets system behaved in a reasonably elastic manner. 4. The measured load distributions generally conform to the nonlinear

pressure distributions proposed by Pell and Turner. 5. It appears that increasing the socket length beyond 2 diam or 3

diam does not increase materially the load supported by bond. 6. The load distribution beneath the bottom of the socket is similar to

a rigid footing resting on an elastic medium. 7. The increased load carrying capacity of the rock sockets proved by

this test resulted in a savings in construction costs of approx 2 million dollars when compared with the original empirical designs.

ACKNOWLEDGMENT

The tests described in this paper were performed as a part of the Sul­fur Dioxide Removal Project for Units 4 and 5 at the Four Corners Power Plant. Mr. Ed Lewis of Arizona Public Service Company was the Project Manager for the participating utilities. United Engineers and Construc­tors, Inc. of Philadelphia, Pa., was the principal engineer and contractor. Sergent, Hauskins, and Beckwith of Phoenix, Ariz., performed the field and lab soil investigations and conducted the load tests. The cooperation and help of Mr. Lewis and Mssrs. Sandy Simko and Everette Chartier of his staff are gratefully acknowledged together with that of Mr. George Beckwith, principal in the firm of Sergent, Hauskins, and Beckwith, and Mssrs. Gino Papola, Jack Moore, and Denny Mohan of United Engineers and Constructors. We would also like to thank Prof. Norbet Schmidt of the University of Missouri-Rolla and Dr. Dennis Duffy of Arizona State University for their assistance in reviewing this manuscript.

APPENDIX I.—-BIBLIOGRAPHY

Aufmuth, R. E., "A Systematic Determination of Engineering Criteria for Rock," Technical Report M-79, U.S. Army Construction Engineering Research Laboratory, 1974.

Christian, J. T., and Carrier, W. D., "Janbu Bjerrum and Kjaernsli's Chart Rein­terpreted," Canadian Geotechnical Journal, Vol. 15, 1978.

Coates, D. F., "Rock Mechanics Principles," Mines Branch Monograph 874, De­partment of Energy, Mines and Resources, Ottawa, Canada, 1970.

Coon, R. F., and Merritt, A. H., "Predicting in Situ Modulus of Deformation Using Rock Quality Indexes," ASTM Special Technical Publication 477, American Society for Testing and Materials, Philadelphia, Pa.

Donald, I. B., Chiu, H. K., and Sloan, S. W., "Theoretical Analysis of Rock Sock­eted Piles," International Conference on Structural Foundations on Rock, Sydney, Australia, 1980.

Freeman, C. F., Klajnerman, D., and Prasad, G. D., "Design of Deep Rock Sock­eted Caissons into Shale Rock," Canadian Geotechnical Journal, Vol. 9, No. 1, 1972.

533

J. Geotech. Engrg. 1983.109:525-535.

Dow

nloa

ded

from

asc

elib

rary

.org

by

IZM

IR Y

UK

SEK

TE

KN

OL

OJI

EN

STIT

USU

on

03/0

7/15

. Cop

yrig

ht A

SCE

. For

per

sona

l use

onl

y; a

ll ri

ghts

res

erve

d.

Gill, S. A., "Design and Construction of Rock Caissons," International Conference on Structural Foundations on Rock, Sydney, Australia, 1980.

Horvath, R. G., and Kinney, T. C , "Shaft Resistance of Rock-Socketed Drilled Piers," Symposium on Deep Foundations, American Society of Civil Engineers, New York, N.Y., 1980, p. 182.

Koutsoftas, D. C , "Caissons Socketed in Sound Mica Schist," Journal of the Geo-technical Engineering Division, ASCE, Vol. 107, No. GT6, June, 1981, pp. 743-757.

Ladanyi, B., discussion of "Friction and End Bearing Tests on Bedrock for High Capacity Socket Design," Canadian Geotechnical Journal, Toronto, Ontario, Can­ada, Vol. 14, No. 1, 1976, pp. 153-155.

Pells, P. J. N., and Turner, R. M., "Elastic Solutions for the Design and Analysis of Rock-Socketed Piles," Canadian Geotechnical Journal, Vol. 16, No. 3, 1979.

Poulos, E. B., and Davis, E. H., "Pile Foundation Design & Analysis," John Wiley and Sons, Inc., 1980.

Rosenberg, P., and Joumeaux, N. L., "Friction and End Bearing Tests on Bed­rock for High Capacity Socket Design," Canadian Geotechnical Journal, Vol. 13, No. 3, 1976.

Rowe, R. K., and Pells, P. J. N., "A Theoretical Study of Pile-Rock Behavior," International Conference on Structural Foundations on Rock, Sydney, Australia, 1980.

"Subsurface Investigation for Design and Construction of Foundations of Build­ings," ASCE-Manuals and Reports of Engineering Practice No. 56, American So­ciety of Civil Engineers, 1976.

Thorburn, S., "Large Diameter Piles Founded on Bedrock," Proceedings of Sym­posium on Large Bored Piles, The Institution of Civil Engineers, London, England, 1966.

Williams, A. F., and Pells, P. J. N., "The Design of Side Resistance Rock Sockets in Sedimentary Rock," Sydney University, Civil Engineering Laboratories, Syd­ney, Australia, 1979.

Williams, A. F., Johnson, I. W., and Donald, I. B., "The Design of Rock-Socketed Piers in Weak Rock," International Conference on Structural Foundations on Rock, Sydney, Australia, 1980.

Williams, A. F., "Principles of Side Resistance Development in Rock Socketed Piles," Third Australia-New Zealand Conference on Geomechanics, Vol. I., Wellington, New Zealand, 1980.

Williams, A. F., and Ervin, M. C , "The Design and Performance of Cast In Situ Piles in Extensively Jointed Silurian Mudstone," Third Australia-New Zealand Con­ference on Geomechanics, Vol. 1, Wellington, New Zealand, 1980.

Wilson, L. C , "Tests on Bored and Driven Piles in Cretaceous Mudstone at Port Elizabeth, South Africa," Piles in Weak Rock, The Institution of Civil Engineers, London, England, 1977.

APPENDIX II.—NOTATION

The following symbols are used in this paper:

534

J. Geotech. Engrg. 1983.109:525-535.

Dow

nloa

ded

from

asc

elib

rary

.org

by

IZM

IR Y

UK

SEK

TE

KN

OL

OJI

EN

STIT

USU

on

03/0

7/15

. Cop

yrig

ht A

SCE

. For

per

sona

l use

onl

y; a

ll ri

ghts

res

erve

d.

a = socket radius; ER = rock modu lus ;

e = settlement; Ip = influence factor; and V = load.

535

J. Geotech. Engrg. 1983.109:525-535.

Dow

nloa

ded

from

asc

elib

rary

.org

by

IZM

IR Y

UK

SEK

TE

KN

OL

OJI

EN

STIT

USU

on

03/0

7/15

. Cop

yrig

ht A

SCE

. For

per

sona

l use

onl

y; a

ll ri

ghts

res

erve

d.